Saturday, June 9, 2007

RE: AISI 12L14 Steel

Thanks Bob.
 
At this time I am checking an item that calls for material diams up to about 3", but I'll try some suppliers directly.
 
Thor
-----Original Message-----
From: Robert Kazanjy [mailto:rkazanjy@gmail.com]
Sent: June 8, 2007 10:06 PM
To: seaint@seaint.org
Subject: Re: AISI 12L14 Steel

Thor-

I'm sure there are properties avaible for larger diameters because larger diameter material is available to purchase.

McMaster has 12L14 per ASTM A108 up to 4"

http://www.pmtsco.com/12L14CD.HTM

has cold drawn 12L14 up to 5 1/2" diameter 

so if you were to buy material from either one of these suppliers I'm sure they could provide chemical & physical certs for the material

I could not find an online source for the mechanical properties for 12L14 in sizes larger than 38mm (1.5")

Unless someone else has them I think you're looking at trip to the library or offering to purchase the info from a material supplier or producer.

cheers
Bob


On 6/8/07, Thor Tandy <vicpeng@telus.net> wrote:
AISI 12L14 Steel

I have data for drawn diams 19 to 38mm.  Does any one know if there is are available properties for greater diams?

Thanks

Thor A. Tandy P.Eng, MIStructE, Struct Eng
Victoria, BC
vicpeng@telus.net

Friday, June 8, 2007

Re: AISI 12L14 Steel

Thor-

I'm sure there are properties avaible for larger diameters because larger diameter material is available to purchase.

McMaster has 12L14 per ASTM A108 up to 4"

http://www.pmtsco.com/12L14CD.HTM

has cold drawn 12L14 up to 5 1/2" diameter 

so if you were to buy material from either one of these suppliers I'm sure they could provide chemical & physical certs for the material

I could not find an online source for the mechanical properties for 12L14 in sizes larger than 38mm (1.5")

Unless someone else has them I think you're looking at trip to the library or offering to purchase the info from a material supplier or producer.

cheers
Bob


On 6/8/07, Thor Tandy <vicpeng@telus.net> wrote:
AISI 12L14 Steel

I have data for drawn diams 19 to 38mm.  Does any one know if there is are
available properties for greater diams?

Thanks

Thor A. Tandy P.Eng, MIStructE, Struct Eng
Victoria, BC
vicpeng@telus.net


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AISI 12L14 Steel

AISI 12L14 Steel

I have data for drawn diams 19 to 38mm. Does any one know if there is are
available properties for greater diams?

Thanks

Thor A. Tandy P.Eng, MIStructE, Struct Eng
Victoria, BC
vicpeng@telus.net


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RE: super flat floors

            That’s o.k. as long as you are working with a reputable specialty contractor like Kalman floors (http://www.kalmanfloor.com/) and understand that you will have to modify some of your typical foundation details so that they can allow for the expansion of the slab before the slab shrinks back to normal size.

 

Matthew Stuart

Structural Department

Manalapan

Extension 1283

 

-----Original Message-----
From: Michelle Motchos [mailto:mmotchos@sw-sc.com]
Sent:
Friday, June 08, 2007 2:14 PM
To: seaint@seaint.org
Subject: RE: super flat floors

 

How about a type K or other shrinkage compensating mix with mild reinforcement? I saw a presentation by Larry Valentine of ShrinkageComp Plus last year that was interesting on the topic (www.shrinkagecomp.com    704-785-0741)

 

 

Michelle Motchos, PE
Stevens & Wilkinson of South Carolina, Inc.

Columbia, SC


From: Dave Handy [mailto:dhandy@trg.ca]
Sent: Friday, June 08, 2007 11:15 AM
To: seaint@seaint.org
Subject: super flat floors

 

Good morning all:

Does anybody out there have experience with super flat slab on grade floors. In our case we have an area of about 200' x 300' with racking height of about 32'. We have spec'd a concrete strength of 25MPa (28 days) with steel fibres, 1 1/2" aggregate and a w/c of 0.55. We are not experts in this fields and as a result are telling the people building it the specs we require and are basically putting the ball in their court. Initially we spec'd a 30MPa concrete with a 0.45+- 0.02 w/c ratio but the concrete supplier said that this mix would crack too much. He tells us that with the cement that they use and the low w/c ratio the strengths would be in the 40+ MPa range and would cause a lot of shrinkage. We are obviously trying to reduce shrinkage and thought that the low w/c ratio along with a requirement for the design slump to be 1 1/2" with super-p being added to increase slump to the 5" range would produce the least shrinkage. We are being told that a w/c ratio of 0.55 or even 0.6 is preferred by the concrete supplier.  The concrete supplier says he takes no responsibility for the mix and has admitted that we can't get guidance from others in different locales because the ingredients are of differing qualities and chemistry than what we deal with in our area.  I have an article from concrete international that recommended 4000 psi with the 0.45+- 0.02 w/c ratio. The concrete supplier has said that he has heard others in the US swear by this mix however he said that if he produces this concrete with his ingredients the strength is far too high and as a result there is more shrinkage cracking.  There does some to be some consensus in what he is saying based upon projects that have been done in our area. Water reducers which I previously thought would reduce shrinkage are actually increasing shrinkage??(according to ACI 360R-06) Why use them at all then?

 

So...

 

What are the key elements that you would use to spec the concrete. We need a final strength in the 4000 psi (28 MPa) range. There is only PC type 10 and blast furnace slag as the cementitious components, the aggregate is limestone. There are no post-tensioned slabs done in our area. The floor people want to let the floor crack and will fill the joints as long in the future as practically possible..perhaps 6 months..with epoxy.

 

any thoughts

 

thanks

Dave

 

 

 

Re: super flat floors

I believe you will find that the experts on these floors in the U.S. is
Kalmar Floors.

Stan Scholl, P.E.
Laguna Beach, CA

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Re: Code errata

Assuming that you did not do your CADD work yourself and used a CADD
operator, would such a CADD operator work for you for free? I doubt it.

In a similar manner, while the technical content of the codes are done by
volunteers that participate for free (although some code committees do
reimburse for some travel expenses...I believe that AISC does), the fact
still remains that the actual document is still put together (i.e.
editted, "layed out" in a page layout publication, etc by paid staff at
organizations. In addition, like it or not, for many of these
organizations, the codes are one of their major sources of income that
allows those organizations to function (the other main source typically is
membership dues...but how many people here are members of various
organizations).

I will use the example that I have used in the past. Considering that it
likely costs you less than a dollar a drawing to print a those drawings
for a client, if you are so convinced that various organizations should
pass on such costs to you to use their intellectual property, why should
use expect a client to pay you significantly more than something less than
a dollar per page of your drawings? And considering it would cost you
even less to PDF those drawings, why would should a client pay you even
than much? The point is that you are not paying for a stack of papers
glued together...if that were the case, then I could go to my local
Staples and buy a stack of paper to sell you for $100. You are paying for
the intellectual property that in contains on those sheets of paper. And
no matter how those organizations created that intellectual property (i.e.
whether it cost them money or not), they are entitle to compensation for
it and have every right to charge what they want.

If you don't like it, then write your own code and get the local
jurisdiction (and others) to adopt it. ;-) To be more serious, it costs
money to produce a code, whether there are volunteers doing it or not. If
you think the free NEHRP provisions cost nothing to produce, then you are
extremely naive at best but maybe delusional. It just so happens that you
don't "see" that cost as it is paid for by making use of tax dollars.

As to your question about the errata, I don't know for sure. If it is
purely an editorial correction/change (i.e. does not change the meaning),
then I would assume that no action is required for it to be "officially"
used. If not (i.e. it does change the meaning), then I would suppose it
is dependent on how each jurisdiction considers it unless there is some
established principle. It is possible that if it fixes something that was
a typo even if it does change the meaning, then it may not need to be
approved as the code committee intended it to be are the errata is showing
it rather than what the typo showed. If the change is to fix something
that the code committee screwed up (i.e. they made a technical mistake),
then I believe that is not typically considered an errata item but rather
a change that needs to be made through the formal code process. Errata is
supposed to only be things that did not make it in the final "printing"
(either hard copy or electronic copy) as was intended, voted upon and
approved by the committee.

Regards,

Scott
Adrian, MI

On Fri, 8 Jun 2007, David Merrick wrote:

> Are errata lists part of the code? Are they being adopted by the local
> governing jurisdictions? How does errata become law, like the code?
>
> If volunteers write the code, why are hundreds of dollars of fees being
> charged for even down-loading it's electronic file? (no printing costs).
> Each errata could be a complete recopy of the whole code with the
> changed print red-lined dated and with approval identification. I could
> understand paying about $10 for the the cost of a web site.
>
> The authors and web sites for the errata get more recognition than the
> volunteers writing the code.
>
> David Merrick, SE
>
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RE: super flat floors

How about a type K or other shrinkage compensating mix with mild reinforcement? I saw a presentation by Larry Valentine of ShrinkageComp Plus last year that was interesting on the topic (www.shrinkagecomp.com    704-785-0741)

 

 

Michelle Motchos, PE
Stevens & Wilkinson of South Carolina, Inc.

Columbia, SC


From: Dave Handy [mailto:dhandy@trg.ca]
Sent: Friday, June 08, 2007 11:15 AM
To: seaint@seaint.org
Subject: super flat floors

 

Good morning all:

Does anybody out there have experience with super flat slab on grade floors. In our case we have an area of about 200' x 300' with racking height of about 32'. We have spec'd a concrete strength of 25MPa (28 days) with steel fibres, 1 1/2" aggregate and a w/c of 0.55. We are not experts in this fields and as a result are telling the people building it the specs we require and are basically putting the ball in their court. Initially we spec'd a 30MPa concrete with a 0.45+- 0.02 w/c ratio but the concrete supplier said that this mix would crack too much. He tells us that with the cement that they use and the low w/c ratio the strengths would be in the 40+ MPa range and would cause a lot of shrinkage. We are obviously trying to reduce shrinkage and thought that the low w/c ratio along with a requirement for the design slump to be 1 1/2" with super-p being added to increase slump to the 5" range would produce the least shrinkage. We are being told that a w/c ratio of 0.55 or even 0.6 is preferred by the concrete supplier.  The concrete supplier says he takes no responsibility for the mix and has admitted that we can't get guidance from others in different locales because the ingredients are of differing qualities and chemistry than what we deal with in our area.  I have an article from concrete international that recommended 4000 psi with the 0.45+- 0.02 w/c ratio. The concrete supplier has said that he has heard others in the US swear by this mix however he said that if he produces this concrete with his ingredients the strength is far too high and as a result there is more shrinkage cracking.  There does some to be some consensus in what he is saying based upon projects that have been done in our area. Water reducers which I previously thought would reduce shrinkage are actually increasing shrinkage??(according to ACI 360R-06) Why use them at all then?

 

So...

 

What are the key elements that you would use to spec the concrete. We need a final strength in the 4000 psi (28 MPa) range. There is only PC type 10 and blast furnace slag as the cementitious components, the aggregate is limestone. There are no post-tensioned slabs done in our area. The floor people want to let the floor crack and will fill the joints as long in the future as practically possible..perhaps 6 months..with epoxy.

 

any thoughts

 

thanks

Dave

 

 

 

RE: Code errata

Are errata lists being adopted by the local governing jurisdictions?

WOW, that's a good question. I'm going to bring that up on the ICC
Forum.


Bob Garner

-----Original Message-----
From: David Merrick [mailto:MRKGP@winfirst.com]
Sent: Friday, June 08, 2007 10:38 AM
To: SEAINT
Subject: Code errata

Are errata lists part of the code? Are they being adopted by the local
governing jurisdictions? How does errata become law, like the code?

If volunteers write the code, why are hundreds of dollars of fees being
charged for even down-loading it's electronic file? (no printing costs).

Each errata could be a complete recopy of the whole code with the
changed print red-lined dated and with approval identification. I could
understand paying about $10 for the the cost of a web site.

The authors and web sites for the errata get more recognition than the
volunteers writing the code.

David Merrick, SE

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Code errata

Are errata lists part of the code? Are they being adopted by the local
governing jurisdictions? How does errata become law, like the code?

If volunteers write the code, why are hundreds of dollars of fees being
charged for even down-loading it's electronic file? (no printing costs).
Each errata could be a complete recopy of the whole code with the
changed print red-lined dated and with approval identification. I could
understand paying about $10 for the the cost of a web site.

The authors and web sites for the errata get more recognition than the
volunteers writing the code.

David Merrick, SE

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*
* This email was sent to you via Structural Engineers
* Association of Southern California (SEAOSC) server. To
* subscribe (no fee) or UnSubscribe, please go to:
*
*

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*
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Re: super flat floors

Dave,
 
        I recently had a problem where I required zero shrinkage.  A product called out as Eclipse Shrinkage Reducing Adx in the amount of 5.5 kg/meter was proposed by the concrete supplier and found acceptable by the materials experts I spoke with.  Perhapse something similar would help solve your problem.
 
Good luck,
 
H. Daryl Richardson
----- Original Message -----
From: Dave Handy
Sent: Friday, June 08, 2007 9:14 AM
Subject: super flat floors

Good morning all:
Does anybody out there have experience with super flat slab on grade floors. In our case we have an area of about 200' x 300' with racking height of about 32'. We have spec'd a concrete strength of 25MPa (28 days) with steel fibres, 1 1/2" aggregate and a w/c of 0.55. We are not experts in this fields and as a result are telling the people building it the specs we require and are basically putting the ball in their court. Initially we spec'd a 30MPa concrete with a 0.45+- 0.02 w/c ratio but the concrete supplier said that this mix would crack too much. He tells us that with the cement that they use and the low w/c ratio the strengths would be in the 40+ MPa range and would cause a lot of shrinkage. We are obviously trying to reduce shrinkage and thought that the low w/c ratio along with a requirement for the design slump to be 1 1/2" with super-p being added to increase slump to the 5" range would produce the least shrinkage. We are being told that a w/c ratio of 0.55 or even 0.6 is preferred by the concrete supplier.  The concrete supplier says he takes no responsibility for the mix and has admitted that we can't get guidance from others in different locales because the ingredients are of differing qualities and chemistry than what we deal with in our area.  I have an article from concrete international that recommended 4000 psi with the 0.45+- 0.02 w/c ratio. The concrete supplier has said that he has heard others in the US swear by this mix however he said that if he produces this concrete with his ingredients the strength is far too high and as a result there is more shrinkage cracking.  There does some to be some consensus in what he is saying based upon projects that have been done in our area. Water reducers which I previously thought would reduce shrinkage are actually increasing shrinkage??(according to ACI 360R-06) Why use them at all then?
 
So...
 
What are the key elements that you would use to spec the concrete. We need a final strength in the 4000 psi (28 MPa) range. There is only PC type 10 and blast furnace slag as the cementitious components, the aggregate is limestone. There are no post-tensioned slabs done in our area. The floor people want to let the floor crack and will fill the joints as long in the future as practically possible..perhaps 6 months..with epoxy.
 
any thoughts
 
thanks
Dave
 
 
 

Mitigation / Acceptance of Concrete Mix with Reactive Aggregates

We have to make determination on acceptance of a concrete mix with a Petrographic examination (ASTM C295) and Mortar Bar Method (ASTM C1260, expansion between 0.1 to 0.2 %) saying Aggregate slightly reactive. Both results done say slightly reactive and it is engineers call to accept or do further investigation. We are using low Alkali cement and project site not very close to sea or salinity environment. (This is the only issue with the mix)

 

Any suggestion/advice or info is appreciated. One of the mitigation Caltran follows is adding Fly ash to mix. 

 

 

Regards,

 

 

Sanjay Kumar Verma,  P.E.

 

 

super flat floors

Good morning all:
Does anybody out there have experience with super flat slab on grade floors. In our case we have an area of about 200' x 300' with racking height of about 32'. We have spec'd a concrete strength of 25MPa (28 days) with steel fibres, 1 1/2" aggregate and a w/c of 0.55. We are not experts in this fields and as a result are telling the people building it the specs we require and are basically putting the ball in their court. Initially we spec'd a 30MPa concrete with a 0.45+- 0.02 w/c ratio but the concrete supplier said that this mix would crack too much. He tells us that with the cement that they use and the low w/c ratio the strengths would be in the 40+ MPa range and would cause a lot of shrinkage. We are obviously trying to reduce shrinkage and thought that the low w/c ratio along with a requirement for the design slump to be 1 1/2" with super-p being added to increase slump to the 5" range would produce the least shrinkage. We are being told that a w/c ratio of 0.55 or even 0.6 is preferred by the concrete supplier.  The concrete supplier says he takes no responsibility for the mix and has admitted that we can't get guidance from others in different locales because the ingredients are of differing qualities and chemistry than what we deal with in our area.  I have an article from concrete international that recommended 4000 psi with the 0.45+- 0.02 w/c ratio. The concrete supplier has said that he has heard others in the US swear by this mix however he said that if he produces this concrete with his ingredients the strength is far too high and as a result there is more shrinkage cracking.  There does some to be some consensus in what he is saying based upon projects that have been done in our area. Water reducers which I previously thought would reduce shrinkage are actually increasing shrinkage??(according to ACI 360R-06) Why use them at all then?
 
So...
 
What are the key elements that you would use to spec the concrete. We need a final strength in the 4000 psi (28 MPa) range. There is only PC type 10 and blast furnace slag as the cementitious components, the aggregate is limestone. There are no post-tensioned slabs done in our area. The floor people want to let the floor crack and will fill the joints as long in the future as practically possible..perhaps 6 months..with epoxy.
 
any thoughts
 
thanks
Dave
 
 
 

Thursday, June 7, 2007

Unbalanced Snow Loads

 
ASCE 7-05 has provisions for "simply supported prismatic members" and "all other gable roofs."  Obviously, trusses are not prismatic and the later provisions apply to designs of trusses.  But, what load is intended to be used for the design of the structure itself? 
 
If the structure uses a truss I've been using the "all other gable roofs" provisions to match the truss reactions.  However, I was recently told that the "simply supported" provisions are meant for the design of the building, regardless of whether the framing is a rafter or a simply supported truss (for W<=20).  If this is done, the resulting design loads can be considerably different from the truss reactions.
 
I realize that snow doesn't know the difference between a rafter and a truss, but what is the intent of ASCE 7 with this regard?  Should the building be designed for different snow loads than the roof framing?
 

Regards,

 

Eric Tompos, PE, SE

 

RE: Glass railing

It is not that uncommon:
http://www.stairservice.com/glass_systems.php?gclid=CI2Bxc-xy4wCFRf9IgodPS76qA

http://www.juliusblum.com

I like to see it load tested after it is installed.

Regards,
Harold Sprague

>From: Tarek Mokhtar <tarooky@earthlink.net>
>Reply-To: <seaint@seaint.org>
>To: <seaint@seaint.org>
>Subject: Glass railing
>Date: Wed, 6 Jun 2007 18:48:00 -0700
>
>Does anybody know of a way to calc glass panels acting as deck guardrails
>that are supported
>only at the lower two corners by bolts? I have seen them around, just not
>sure how to make them
>work, or if somebody is interested in taking the job, please email me
>privately
>
>Thanx in advance
>--
>
>Tarek Mokhtar, SE
>Laguna Beach, CA

_________________________________________________________________
PC Magazine's 2007 editors' choice for best Web mail—award-winning Windows
Live Hotmail.

http://imagine-windowslive.com/hotmail/?locale=en-us&ocid=TXT_TAGHM_migration_HM_mini_pcmag_0507


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RE: Torquing of bolts by the bolt head

I would recommend that if you in fact need the bolt to be tensioned:

1. Weld a carbon steel retainer on the structural steel. The nut is
quenched and tempered and should not be welded. A carbon steel retainer is
relatively easy to fabricate and can be welded to the structural steel to
hold the nut in place.

2. Use a direct tension indicating washer to determine when the bolt is
properly tensioned.

Regards,
Harold Sprague

>From: "Michael Laplante" <Michael.Laplante@cima.ca>
>Reply-To: <seaint@seaint.org>
>To: <SEAINT@SEAINT.ORG>
>CC: <jacques.caron@cima.ca>
>Subject: Torquing of bolts by the bolt head
>Date: Thu, 7 Jun 2007 15:48:48 -0400
>
>I am connecting several new steel beams into an existing structure.
>Due to the confined space of some of the connections access to the nuts
>of the bolted connections is limited. Therefore nuts are pre-welded to
>one side of the connection and the bolts are inserted by the other side.
>Question? Can the bolts be torqued or pre-tensioned by turning the head
>of the bolt rather than the nut. A washer is installed on the head side
>of the connection. I don't think there is a difference as long as the
>washer is installed behind the head.
>
>Thanks in advance
>
>Mike
>CIMA + Laval
>Structures- Energie
>W: 514-337-2462
>C: 514-229-5340
>F: 450-682-1013
>
>
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Re: How do I put my question in this forum

You just did.
 
 
Will H.
 
On 6/7/07, Sanjay Verma <sverma@p-and-i.com> wrote:

Could you please advice how I can put my question in this forum

 

Regards,

 

 

Sanjay Kumar Verma,  P.E.

 

 


Re: How do I put my question in this forum

You just did it!  What other questions do you have?

Ralph

In a message dated 6/7/07 5:55:42 PM, sverma@p-and-i.com writes:

Could you please advice how I can put my question in this forum
 
Regards,
 
 
Sanjay Kumar Verma,  P.E.
 



**************************************
See what's free at http://www.aol.com.

How do I put my question in this forum

Could you please advice how I can put my question in this forum

 

Regards,

 

 

Sanjay Kumar Verma,  P.E.

 

 

Re: Glass railing Redux

List,

I tried to send two photos I recently took in the new wing of the Phoenix Art Museum, but they bounced -- "too big."  They show 1) a glass railing with NO top rail or continuity, each piece of glass attached to the edge of a concrete slab with 4 big "buttons," and 2) a glass stair railing with a handrail hanging on the side of it, each piece of glass attached to the edge of the concrete stair slab with *2* of those shiny "buttons."  It really looks neat; I wonder how they do it.  It's thick glass, thicker than 1/2" I believe. 

Ralph

In a message dated 6/7/07 5:32:33 PM, Mlcse@aol.com writes:
How wide is the glass between the bolts.  If you check the glass in out-of-plane bending along a yield line through the bolts, you will probably find it doesn't work in flexure.  Yes you need the top rail continuous over three pieces of glass incase one piece breaks, but each individual sheet of cantilevered glass has to be checked for the guardrail loading acting at the top of the glass, and generally you need a piece of glass 4 feet wide to resist the applied loading from my experience for 1/2" thick glass.
 
Michael Cochran
 
In a message dated 6/7/2007 8:19:52 A.M. Pacific Daylight Time, tarooky@earthlink.net writes:

The glass railing will have a top rail, but my understanding is that the top rail is required
over at least three panels so that if one fails, the top rail can still be supported. however,
the glass panel is still cantilevered - kind of - from the corners, with four bolts in each corner

Tarek Mokhtar, SE
Laguna Beach, CA








I don't think this is going to work with just a bolt in the lower corners of the glass, your stresses are going to be way to high.  You need a continuous shoe for the glass to frame into if its a cantilever glass guardrail, and the glass needs to be about 4 feet long in the shoe to work to handle the 200 pount point load at the top of the glass railing (especially if its only 1/2" thick glass).  There are design requirements in the UBC and IBC for glass railings that a lot of people either don't know about or choose to ignore.  I have seen the glass work with a bolt in each corner it you have a continuous rail above the glass to act as the guard rail, and the glass is just an infill panel between the top and bottom continuous rails between vertical steel stanchions.

Michael Cochran SE

-----Original Message-----
From: Tarek Mokhtar
To: seaint@seaint.org
Sent: Wed, 6 Jun 2007 6:48 pm
Subject: Glass railing


Does anybody know of a way to calc glass panels acting as deck guardrails that are supported 
only at the lower two corners by bolts? I have seen them around, just not sure how to make them 
work, or if somebody is interested in taking the job, please email me privately 
 
Thanx in advance 
--  
Tarek Mokhtar, SE 
Laguna Beach, CA 



**************************************
See what's free at http://www.aol.com.

Re: Torquing of bolts by the bolt head

On Jun 7, 2007, at 3:28 PM, Garner, Robert wrote:

> But is it safe to weld nuts (heat-treated steel and all)?
No. It's sometimes done, but the carbon equivalent is way too high
and the welding is very tricky because the fillet lengths are very
short. Tack welds in that sort of steel are prone to crack. There's
also distortion to worry about. A client of mine invariably prefers
drilling and tapping where stripping strength isn't an issue.

Christopher Wright P.E. |"They couldn't hit an elephant at
chrisw@skypoint.com | this distance" (last words of Gen.
.......................................| John Sedgwick, Spotsylvania
1864)
http://www.skypoint.com/~chrisw/

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Re: Glass railing

How wide is the glass between the bolts.  If you check the glass in out-of-plane bending along a yield line through the bolts, you will probably find it doesn't work in flexure.  Yes you need the top rail continuous over three pieces of glass incase one piece breaks, but each individual sheet of cantilevered glass has to be checked for the guardrail loading acting at the top of the glass, and generally you need a piece of glass 4 feet wide to resist the applied loading from my experience for 1/2" thick glass.
 
Michael Cochran
 
In a message dated 6/7/2007 8:19:52 A.M. Pacific Daylight Time, tarooky@earthlink.net writes:
The glass railing will have a top rail, but my understanding is that the top rail is required
over at least three panels so that if one fails, the top rail can still be supported. however,
the glass panel is still cantilevered - kind of - from the corners, with four bolts in each corner

Tarek Mokhtar, SE
Laguna Beach, CA







I don't think this is going to work with just a bolt in the lower corners of the glass, your stresses are going to be way to high.  You need a continuous shoe for the glass to frame into if its a cantilever glass guardrail, and the glass needs to be about 4 feet long in the shoe to work to handle the 200 pount point load at the top of the glass railing (especially if its only 1/2" thick glass).  There are design requirements in the UBC and IBC for glass railings that a lot of people either don't know about or choose to ignore.  I have seen the glass work with a bolt in each corner it you have a continuous rail above the glass to act as the guard rail, and the glass is just an infill panel between the top and bottom continuous rails between vertical steel stanchions.

Michael Cochran SE

-----Original Message-----
From: Tarek Mokhtar
To: seaint@seaint.org
Sent: Wed, 6 Jun 2007 6:48 pm
Subject: Glass railing
Does anybody know of a way to calc glass panels acting as deck guardrails that are supported 
only at the lower two corners by bolts? I have seen them around, just not sure how to make them 
work, or if somebody is interested in taking the job, please email me privately 
 
Thanx in advance 
--  
Tarek Mokhtar, SE 
Laguna Beach, CA 
 
 




See what's free at AOL.com.

RE: ASCE 7-05 Errors

Yes, no set of construction documents is 100% error free and no technical
publication is 100% error free, but 24 pages of errata sounds excessive.
Most constr. docs errors are typos inwhich we can interpret the meaning,
(ie. "wood beam" spelled "wood bem." But a mistake in a formula in a
technical publication can have disasterous results, (i.e. a "greater than"
sign reversed to "less than"). From the postings, it sounds like ASCE
dropped the ball at our expense.

Larry Hauer S.E.


>From: "Mark E. Deardorff" <mdeardorff@burkett-wong.com>
>Reply-To: <seaint@seaint.org>
>To: <seaint@seaint.org>
>Subject: RE: ASCE 7-05 Errors
>Date: Thu, 7 Jun 2007 13:27:12 -0700
>
>Not as bad as AISC's goof on the Seismic Design Manual. It was too
>extensive
>to treat with errata.
>
>Mark E. Deardorff, S.E.
>Structural Engineer
> <http://www.burkett-wong.com/> Burkett & Wong Engineers
>3434 4th Ave
>San Diego, CA 92103
>P 619.299.5550
>F 619.299.9934
>mdeardorff@burkett-wong.com
>
>
>
> _____
>
>From: Garner, Robert [mailto:rgarner@moffattnichol.com]
>Sent: Wednesday, June 06, 2007 2:15 PM
>To: seaint@seaint.org
>Subject: ASCE 7-05 Errors
>
>
>
>Just a tip that the latest errata for ASCE 7-05 is available from
>SEInstitute.org All twenty four pages! In addition, S.K. Ghosh
>published
>some ASCE 7 Seismic Provisions Errata in the April issue of Structural
>Engineer. I don't know if Ghosh's errata was picked up by SEInstitute or
>not.
>
>
>
>These errata are so extensive that I think you can ignore buying ASCE 7-05
>and just download the errata for free.
>
>
>
>I think I'm just going to go ahead and do really sloppy engineering on my
>next project - I can always issue errata, right?
>
>
>
>Bob Garner, S.E.
>
>
>
>R. Garner
>
>Moffatt & Nichol
>
>Tel.: (619) 220-6050
>
>Fax.: (619) 220-6055
>
>e-mail: rgarner@moffattnichol.com
>
>
>
>The information contained in the e-Mail, including any accompanying
>documents or attachments, is from Moffatt & Nichol and is intended only for
>the use of the individual or entity named above, and is privileged and
>confidential. If you are not the intended recipient, be aware that any
>disclosure, dissemination, distribution, copying or use of the contents of
>this message is strictly prohibited. If you received this message in error,
>please notify us.
>
>

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RE: Torquing of bolts by the bolt head

RCSC Specifications for Structural Joints Using ASTM A325 or A490 Bolts, Section 8.2, allows turning of either.  This is for Pretensioned Joints.

But is it safe to weld nuts (heat-treated steel and all)?  See other threads on welding nuts.

 

Bob Garner, S.E.

 


From: Michael Laplante [mailto:Michael.Laplante@cima.ca]
Sent: Thursday, June 07, 2007 12:49 PM
To: SEAINT@SEAINT.ORG
Cc: jacques.caron@cima.ca
Subject: Torquing of bolts by the bolt head

 

I am connecting  several new steel beams into an existing structure.  Due to the confined space of some of the connections access to the nuts of the bolted connections is limited. Therefore  nuts are pre-welded to one side of the connection and the bolts are inserted by the other side. Question? Can the bolts be torqued or pre-tensioned by turning the head of the bolt rather than the nut.  A washer is installed on the head side of the connection.  I don't think there is a difference as long as the washer is installed behind the head. 

 

Thanks in advance

 

Mike

CIMA + Laval

Structures- Energie

W: 514-337-2462

C:  514-229-5340

F:  450-682-1013

 

 


 

 

RE: ASCE 7-05 Errors

Not as bad as AISC's goof on the Seismic Design Manual. It was too extensive to treat with errata.
 
Mark E. Deardorff, S.E.
Structural Engineer
Burkett & Wong Engineers
3434 4th Ave
San Diego, CA 92103
P 619.299.5550
F 619.299.9934
mdeardorff@burkett-wong.com
 


From: Garner, Robert [mailto:rgarner@moffattnichol.com]
Sent: Wednesday, June 06, 2007 2:15 PM
To: seaint@seaint.org
Subject: ASCE 7-05 Errors

Just a tip that the latest errata for ASCE  7-05 is available from SEInstitute.org    All twenty four pages!  In addition, S.K. Ghosh published some ASCE 7 Seismic Provisions Errata in the April issue of Structural Engineer.  I don't know if Ghosh's errata was picked up by SEInstitute or not.

 

These errata are so extensive that I think you can ignore buying ASCE 7-05 and just download the errata for free.

 

I think I'm just going to go ahead and do really sloppy engineering on my next project - I can always issue errata, right?

 

Bob Garner, S.E.

 

R. Garner

Moffatt & Nichol

Tel.:  (619) 220-6050

Fax.: (619) 220-6055

e-mail: rgarner@moffattnichol.com

 

The information contained in the e-Mail, including any accompanying documents or attachments, is from Moffatt & Nichol and is intended only for the use of the individual or entity named above, and is privileged and confidential. If you are not the intended recipient, be aware that any disclosure, dissemination, distribution, copying or use of the contents of this message is strictly prohibited. If you received this message in error, please notify us.

Torquing of bolts by the bolt head

I am connecting  several new steel beams into an existing structure.  Due to the confined space of some of the connections access to the nuts of the bolted connections is limited. Therefore  nuts are pre-welded to one side of the connection and the bolts are inserted by the other side. Question? Can the bolts be torqued or pre-tensioned by turning the head of the bolt rather than the nut.  A washer is installed on the head side of the connection.  I don't think there is a difference as long as the washer is installed behind the head. 
 
Thanks in advance
 
Mike
CIMA + Laval
Structures- Energie
W: 514-337-2462
C:  514-229-5340
F:  450-682-1013
 
 

 
 

Re: ASCE 7-05 Errors

I think they actually balked at the IBC due to the NFPA and the fireman's union. I am for the centralization of codes by reference. So I have to by AISC manual...I do anyway. I have to buy the ACI...every engineer working in concrete should. NDS is cheap. And the ASCE is another code. So say a 600 bucks every 3-4 years (hopefully that extends) you got most of your codes. Doesn't seem so bad, an I'm sure most of your decent size offices buy most of that stuff for you anyway (or at least has a library for all to share).

I did a few projects on the IBC 2001 a few years back. I found that transition fairly easy. I haven't used ASCE-7 yet, but will surely once Jan. comes along. The older versions were well written I thought.

I guess the new ones have lots of TYPOS....the UBC 97 had the same problems, and issued Eratta after Eratta and even flip flopped on code issues (not a mere typo) on certain things. It's is unfortunate.

Sounds like to me, there needs to be a professional editing company involved (or a new one) for syntax issues, but the technical stuff, I can't see it being done by anyone other than in-house.

-g

On 6/7/07, Donald Bruckman <bruckmandesign@verizon.net> wrote:

I just started going through the IBC-2006 in anticipation of its use here in California next year.  So far, I am not impressed.  AT ALL.  From what I've looked at so far, the IBC is a rather sloppy document and surprisingly, in many areas, much more lax than UBC-97 ….  I haven't even gotten into the structural stuff and I see all kinds of poor phrasing and awkward syntax, errors of omission, unreferenced requirements and other stuff that will just make my life a living hell trying to interpret in unison with a B.O..  No wonder the State of California balked at this document the first go-around. 

 

I also see why you guys were complaining about cost.  Seems the IBC simply references an alphabet soup of other documents and simply says, "…shall comply with ASCE-7…" or AITC or ASTM or whatever.  Meanwhile, the State of CA has two pdf documents totaling over 300 pages of amendments.  Entire chapters are omitted and re-written.

 

Yikes.

 


From: Garner, Robert [mailto:rgarner@moffattnichol.com]
Sent: Thursday, June 07, 2007 10:22 AM
To: seaint@seaint.org
Subject: ASCE 7-05 Errors

 

ASCE Page 83, Section 7.7.1, Last Sentence: "This density shall also be used to determine h sub b by dividing p sub s by lambda."

 

p sub s is the sloped roof snow load.  They should also include the flat roof snow load, p sub f.

 

This apparently got by all those errata.

 

With all due respect, I still consider the errata excessive for a Design Code document.  And, yes, we strive for 100% accuracy in our office.  Errors are not treated lightly here.

 

 

Sincerely,

 

Bob Garner, S.E.

 

R. Garner

Moffatt & Nichol

Tel.:  (619) 220-6050

Fax.: (619) 220-6055

e-mail: rgarner@moffattnichol.com

 

The information contained in the e-Mail, including any accompanying documents or attachments, is from Moffatt & Nichol and is intended only for the use of the individual or entity named above, and is privileged and confidential. If you are not the intended recipient, be aware that any disclosure, dissemination, distribution, copying or use of the contents of this message is strictly prohibited. If you received this message in error, please notify us.



--
-gm

RE: ASCE 7-05 Errors

I just started going through the IBC-2006 in anticipation of its use here in California next year.  So far, I am not impressed.  AT ALL.  From what I’ve looked at so far, the IBC is a rather sloppy document and surprisingly, in many areas, much more lax than UBC-97 ….  I haven’t even gotten into the structural stuff and I see all kinds of poor phrasing and awkward syntax, errors of omission, unreferenced requirements and other stuff that will just make my life a living hell trying to interpret in unison with a B.O..  No wonder the State of California balked at this document the first go-around. 

 

I also see why you guys were complaining about cost.  Seems the IBC simply references an alphabet soup of other documents and simply says, “…shall comply with ASCE-7…” or AITC or ASTM or whatever.  Meanwhile, the State of CA has two pdf documents totaling over 300 pages of amendments.  Entire chapters are omitted and re-written.

 

Yikes.

 


From: Garner, Robert [mailto:rgarner@moffattnichol.com]
Sent: Thursday, June 07, 2007 10:22 AM
To: seaint@seaint.org
Subject: ASCE 7-05 Errors

 

ASCE Page 83, Section 7.7.1, Last Sentence: "This density shall also be used to determine h sub b by dividing p sub s by lambda."

 

p sub s is the sloped roof snow load.  They should also include the flat roof snow load, p sub f.

 

This apparently got by all those errata.

 

With all due respect, I still consider the errata excessive for a Design Code document.  And, yes, we strive for 100% accuracy in our office.  Errors are not treated lightly here.

 

 

Sincerely,

 

Bob Garner, S.E.

 

R. Garner

Moffatt & Nichol

Tel.:  (619) 220-6050

Fax.: (619) 220-6055

e-mail: rgarner@moffattnichol.com

 

RE: steel - vibration analysis

Thanks for the contact, Will.  I got in touch with Allen Adams at RAM/Bentley and he was very helpful in clarifying the dilemma I encountered.  After discussion with Dr. Murray who developed the FloorVibe program, everyone agreed that the AISC Design Guide is somewhat ambiguous when it comes to irregular framing/bays.  Additionally, the issue I brought up was WRT a “mezzanine.”  Traditionally, a mezzanine is a partial floor between two complete floors.  In my situation, the mezzanine is considered a floor, and the vibration analysis became much more realistic.  Many thanks to Mr. Adams, and others for their input.

 

David A. Topete, SE

 


From: Will Haynes [mailto:gtg740p@gmail.com]
Sent: Friday, June 01, 2007 6:56 AM
To: seaint@seaint.org
Subject: Re: steel - vibration analysis

 

Allen Adams at RAM can help you if you are getting different answers between the Design Guide 11 example and Floorvibe via RAM. You should get the same.


Will Haynes




On 5/31/07, Josh Plummer <josh.plummer@cox.net> wrote:

David -

 

I've used the FloorVibe software, but only the share ware version that AISC had on their website for awhile.  Not sure if they have a better / commercial version or not. Anyway, the problem that I see with this software is that everything has to be nice and orthogonal.  I didn't see a way to enter in the types of skewed project grids that I really see on design projects.  My impression was that it was even more limited than that, but it's been awhile so I can't quite remember what seemed so limiting about it. 

 

I don't have any experience with RAM, but keep in mind that the values handed to FloorVibe from RAM would have to be simplified and approximated to fit into the very rigid constraints of what FloorVibe is allowed to do.  The assumptions and simplifications that RAM makes may not be all that good.  If the guys at RAM really new how to to floor vibrations, then I'd think that they would just do the calculation themselves?  This may not be simple for the average engineer who does this calculations once every few years.  However, it's shouldn't be all that difficult for a development team whose job it is to get it right once when they're programming it. 

 

Disclaimer: I work for a structural engeineering software company (RISA Technologies) that competes against RAM in the Floor design market and which also brags about our ability to do DG-11 floor vibration calculations.  I'm ridiculously busy these days, but if you were to e-mail me a good description of your floor system or some drawings and such, then I can try to get one of our engineers to enter this into our Floor program to demonstrate our ability to do the DG-11 calculations accurately.

 

Sincerely,

 

 

Josh Plummer, SE

RISA Technologies

 


From: David Topete [mailto:dtopete@gfdseng.com]
Sent: Thursday, May 31, 2007 4:44 PM
To: seaint@seaint.org
Subject: steel - vibration analysis

Checking vibration on steel framing for a mezzanine, I got 2 similar answers by following example 4.4 of AISC Design Guide 11 (Designing for Floor Vibrations) and the Floor Framing software from the www.AISC.org/steeltools downloadable software.  While those answers were similar, the analysis provided by FloorVibe via Ram Structural System indicated members being increased about 60% (conservatively).  Has anyone else encountered such a discrepancy?

 

David A. Topete, SE

Structural Engineer

 

GFDS Engineers

543 Howard St., First Floor

San Francisco, CA 94105

v : (415) 512-1301 x21

f : (415) 512-1302

dtopete@gfdseng.com

www.gfdseng.com