Structural Engineering: 2008-01-06

Friday, January 11, 2008

RE: resisting moment

For wind:

For IBC 2003, try section 1609.1.3

For IBC 2006, try section 1605.3.2

Both sections say that the 2/3 D is only when using the Alt. Basic load combo’s

If using the basic load combo’s just follow the load multipliers

For seismic:

Just follow the load multipliers for either set of load combo’s

Jason

-----Original Message-----
From: DA ENGINEERING [mailto:dnae@cox.net]
Sent: Friday, January 11, 2008 1:43 PM
To: seaint@seaint.org
Subject: resisting moment

Just wonder where in the code resisting load for uplift for wood shear wall

for wind and seismic

I see this equation in the book SEA #2 example (0.6- 0.14Sds)Mr

I know for wind stays the same 2/3Mr

but I can find in the code

Thanks

Dave A.

Re: Difference between SCBF OCBF

Unless the jurisdiction specifically states that you must use SCBF for an essential facility, it appears that you could use either. Remember that the R values are dramatically different, as are the Cd deflection amplification factors for either system. Also, check IBC 1613.5.6, based on the site spectral acceleration values.

On Jan 11, 2008 9:21 AM, Sandman <sandman21@gmail.com> wrote:

I am currently working on the design of a building with an Occupancy category of IV in the Southern California area. The building meets the requirements for OCBF Seismic Design Cat. D and is under 35'. I was wondering if there are any other requirements, since the building is an essential facility, which would require me to use SCBF?

--
David Topete, SE

Re: Hollow Core Plank Recommendation

Rich,
I did a precast plank job for office loading, and I ended up with a 12" light precast plank with a minimum 3-1/2" topping. It pushed the limits of the design. I remember vibration was a concern because we could make the units work for the expected loading. good luck.

On Jan 10, 2008 5:40 PM, Rich Lewis <seaint04@lewisengineering.com> wrote:

It has been a long time since I did a precast plank job. I will need to get old paper drawings out for details because it was pre-CAD. I now have a plank project for a dormitory style building. I'm trying to remember the span limits for plank. I know the manufacturer's have load tables, but I'm trying to remember the practical experience limits and it is not coming back to me.

I have a 35 ft. span. I'm thinking that it should most likely be a 10" plank, even though I can find load tables for 8" plank that say they work.

From practical experience, what should I limit spans to for 8", 10" and 12" plank, both topped and untopped?

Thanks for your insight.

Rich

--
David Topete, SE

resisting moment

Just wonder where in the code resisting load for uplift for wood shear wall

for wind and seismic

I see this equation in the book SEA #2 example (0.6- 0.14Sds)Mr

I know for wind stays the same 2/3Mr

but I can find in the code

Thanks

Dave A.

RE: Dual - System ( frame+wall)

I think YH is referring to frame/shear wall in same plane no?

Thor A. Tandy P.Eng, MIStructE, Struct Eng
Victoria, BC
Canada
vicpeng@telus.net

-----Original Message-----
From: Gerard Madden, SE [mailto:gmse4603@gmail.com]
Sent: Friday, January 11, 2008 9:14 AM
To: seaint@seaint.org
Subject: Re: Dual - System ( frame+wall)

Jordan is correct.

The moment frames in a "dual system" are intended as a "BACK UP" system to help out a stiffer lateral system elsewhere in the building. The moment frame portion is not supposed to be able to handle 100% of the lateral force considering the stiffer system as 100% completely failed. It's intent is to take up the slack, if any, as the stiff elements begins to deflect excessively ( i.e. yield)

Code requirements (well in the UBC) have use 25% of the total lateral force as the baseshear design value for the moment frames.

They are typically placed at the perimeter, can help lesses torsional irregularities and drift when the stiffer system is primarily in the core.

-gm

On Jan 11, 2008 5:02 AM, Jordan Truesdell, PE <seaint2@truesdellengineering.com> wrote:

Is this borne out by testing? Aside from the fact that this seems to be a cost-inefficient way to build (walls enough to carry most of the load plus a fully redundant steel moment frame), in a quasi-static case, the walls would be taking a multiple of the load seen by the frame (due to the response factor design methodology), resulting in a sudden collapse as the cmu walls failed.
Jordan
y.hamida wrote:

yes: the walls will take load as their rigidity but when the stresses in the walls reinforcement reach mor than

the yielding point the walls will crack in the plastic phase, and will loose their rigidty then the frames will

intervene and take the entire base shear because the frames already design to resist the entire shear.

                  Dr.hamida

----- Original Message -----

From: Thor Tandy

To: seaint@seaint.org

Sent: Thursday, January 10, 2008 11:56 AM

Subject: RE: Dual - System ( frame+wall)

?

... surely the stiffer element will take the load up before the frame deforms enough to carry loads ...? Or is this suggesting that the base shear of a dual system is greater for frames than for shear walls ...?

Thor A. Tandy P.Eng, MIStructE, Struct Eng
Victoria, BC
Canada
vicpeng@telus.net

-----Original Message-----
From: y.hamida [mailto:y.hamida@scs-net.org]
Sent: Thursday, January 10, 2008 11:36 AM
To: seaint@seaint.org
Subject: Dual - System ( frame+wall)

                                                                  Dual - System

        A - code sayes if you have shear walls and ordinary frames (   beams+columns )

                              to resist earthquake
                          You can neglect the ordinary frames and resist the entire base shear by shear walls

                            and designe the ordinary frames just for axial loads.

                           .

       B - M y (ph.d) research sayes;

              if you have  moment resisting frames and walls( reinforce concrete or masonary  )

              to resist earthquake You can neglect the walls and resist the entire base shear

               by  moment resisting frames  and the walls design just for axial loads



             and minimum reinforcement for concrete walls and conect the masonary walls to

                the floor by dowels  .

                        Dr .hamida
******* ****** ******* ******** ******* ******* ******* *** * Read list FAQ at: http://www.seaint.org/list_FAQ.asp * * This email was sent to you via Structural Engineers * Association of Southern California (SEAOSC) server. To * subscribe (no fee) or UnSubscribe, please go to: * * http://www.seaint.org/sealist1.asp * * Questions to seaint-ad@seaint.org. Remember, any email you * send to the list is public domain and may be re-posted * without your permission. Make sure you visit our web * site at: http://www.seaint.org ******* ****** ****** ****** ******* ****** ****** ********

--
-gm

Difference between SCBF OCBF

I am currently working on the design of a building with an Occupancy category of IV in the Southern California area. The building meets the requirements for OCBF Seismic Design Cat. D and is under 35'. I was wondering if there are any other requirements, since the building is an essential facility, which would require me to use SCBF?

Re: Dual - System ( frame+wall)

On Jan 11, 2008 5:02 AM, Jordan Truesdell, PE <seaint2@truesdellengineering.com> wrote:

Is this borne out by testing? Aside from the fact that this seems to be a cost-inefficient way to build (walls enough to carry most of the load plus a fully redundant steel moment frame), in a quasi-static case, the walls would be taking a multiple of the load seen by the frame (due to the response factor design methodology), resulting in a sudden collapse as the cmu walls failed.
Jordan
y.hamida wrote:

yes: the walls will take load as their rigidity but when the stresses in the walls reinforcement reach mor than

the yielding point the walls will crack in the plastic phase, and will loose their rigidty then the frames will

intervene and take the entire base shear because the frames already design to resist the entire shear.

                  Dr.hamida

----- Original Message -----

From: Thor Tandy

To: seaint@seaint.org

Sent: Thursday, January 10, 2008 11:56 AM

Subject: RE: Dual - System ( frame+wall)

?

... surely the stiffer element will take the load up before the frame deforms enough to carry loads ...? Or is this suggesting that the base shear of a dual system is greater for frames than for shear walls ...?

Thor A. Tandy P.Eng, MIStructE, Struct Eng
Victoria, BC
Canada
vicpeng@telus.net

-----Original Message-----
From: y.hamida [mailto:y.hamida@scs-net.org]
Sent: Thursday, January 10, 2008 11:36 AM
To: seaint@seaint.org
Subject: Dual - System ( frame+wall)

                                                                  Dual - System

        A - code sayes if you have shear walls and ordinary frames (   beams+columns )

                              to resist earthquake
                          You can neglect the ordinary frames and resist the entire base shear by shear walls

                            and designe the ordinary frames just for axial loads.

                           .

       B - M y (ph.d) research sayes;

              if you have  moment resisting frames and walls( reinforce concrete or masonary  )

              to resist earthquake You can neglect the walls and resist the entire base shear

               by  moment resisting frames  and the walls design just for axial loads



             and minimum reinforcement for concrete walls and conect the masonary walls to

                the floor by dowels  .

                        Dr .hamida
******* ****** ******* ******** ******* ******* ******* *** * Read list FAQ at: http://www.seaint.org/list_FAQ.asp * * This email was sent to you via Structural Engineers * Association of Southern California (SEAOSC) server. To * subscribe (no fee) or UnSubscribe, please go to: * * http://www.seaint.org/sealist1.asp * * Questions to seaint-ad@seaint.org. Remember, any email you * send to the list is public domain and may be re-posted * without your permission. Make sure you visit our web * site at: http://www.seaint.org ******* ****** ****** ****** ******* ****** ****** ********

--
-gm

Re: Dual - System ( frame+wall)

Is this borne out by testing? Aside from the fact that this seems to be a cost-inefficient way to build (walls enough to carry most of the load plus a fully redundant steel moment frame), in a quasi-static case, the walls would be taking a multiple of the load seen by the frame (due to the response factor design methodology), resulting in a sudden collapse as the cmu walls failed.

Jordan

y.hamida wrote:

yes: the walls will take load as their rigidity but when the stresses in the walls reinforcement reach mor than

the yielding point the walls will crack in the plastic phase, and will loose their rigidty then the frames will

intervene and take the entire base shear because the frames already design to resist the entire shear.

                  Dr.hamida

----- Original Message -----

From: Thor Tandy

To: seaint@seaint.org

Sent: Thursday, January 10, 2008 11:56 AM

Subject: RE: Dual - System ( frame+wall)

?

... surely the stiffer element will take the load up before the frame deforms enough to carry loads ...? Or is this suggesting that the base shear of a dual system is greater for frames than for shear walls ...?

Thor A. Tandy P.Eng, MIStructE, Struct Eng
Victoria, BC
Canada
vicpeng@telus.net

-----Original Message-----
From: y.hamida [mailto:y.hamida@scs-net.org]
Sent: Thursday, January 10, 2008 11:36 AM
To: seaint@seaint.org
Subject: Dual - System ( frame+wall)

                                                                  Dual - System

        A - code sayes if you have shear walls and ordinary frames (   beams+columns )

                              to resist earthquake
                          You can neglect the ordinary frames and resist the entire base shear by shear walls

                            and designe the ordinary frames just for axial loads.

                           .

       B - M y (ph.d) research sayes;

              if you have  moment resisting frames and walls( reinforce concrete or masonary  )

              to resist earthquake You can neglect the walls and resist the entire base shear

               by  moment resisting frames  and the walls design just for axial loads



             and minimum reinforcement for concrete walls and conect the masonary walls to

                the floor by dowels  .

                        Dr .hamida

Thursday, January 10, 2008

RE: Lighter Loads in IBC vs. UBC

Chris,

The few wind and seimic designs I have done under the 06 IBC basically confirms your belief in that the lateral loading is similar or slightly less than the '97 UBC for wood framed buildings- This applies to the Ventura County, California area.

Larry Hauer S.E.

> Date: Wed, 9 Jan 2008 12:29:16 -0800
> From: chris@jdwylieengineering.com
> To: seaint@seaint.org
> Subject: Lighter Loads in IBC vs. UBC
>
> So I think I need a sanity check. Last time I asked a question here I
> got blasted at first and then wound up getting some very useful
> information. So I suppose I'll give it another shot!
>
> We're working on making the transition from 1997 UBC to 2006 IBC.
> While the wind loads are more complicated, I'm finding that as long as
> topography isn't a factor, even the highest load (that of the end
> conditions) is the same or lower than what we were using in the UBC.
> So we can simplify and use the "A" load across the whole projected
> area, and still generate Wind loads that are lower than what we were
> designing for before.
>
> Similarly, when I go through the USGS software and get values for S1
> and Ss, then put that through the equations in ASCE7-05, I come up
> with a Base Shear equation that is a fair bit lower. Where a one
> story home in the Bay Area would have previously had a V of say 0.14W,
> the new code puts it at 0.108W.
>
> I can send more details of how I'm coming up with these figures, but I
> just wanted to start by finding out if other folks are seeing the same
> thing.
>
> Thanks in advance.
>
> Chris Slater, PE
>
> ******* ****** ******* ******** ******* ******* ******* ***
> * Read list FAQ at: http://www.seaint.org/list_FAQ.asp
> *
> * This email was sent to you via Structural Engineers
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> * subscribe (no fee) or UnSubscribe, please go to:
> *
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Make distant family not so distant with Windows Vista® + Windows Live™. Start now!

Hollow Core Plank Recommendation

It has been a long time since I did a precast plank job. I will need to get old paper drawings out for details because it was pre-CAD. I now have a plank project for a dormitory style building. I’m trying to remember the span limits for plank. I know the manufacturer’s have load tables, but I’m trying to remember the practical experience limits and it is not coming back to me.

I have a 35 ft. span. I’m thinking that it should most likely be a 10” plank, even though I can find load tables for 8” plank that say they work.

From practical experience, what should I limit spans to for 8”, 10” and 12” plank, both topped and untopped?

Thanks for your insight.

Rich

Re: Dual - System ( frame+wall)

yes: the walls will take load as their rigidity but when the stresses in the walls reinforcement reach mor than

the yielding point the walls will crack in the plastic phase, and will loose their rigidty then the frames will

intervene and take the entire base shear because the frames already design to resist the entire shear.

Dr.hamida

----- Original Message -----

From: Thor Tandy

To: seaint@seaint.org

Sent: Thursday, January 10, 2008 11:56 AM

Subject: RE: Dual - System ( frame+wall)

?

... surely the stiffer element will take the load up before the frame deforms enough to carry loads ...? Or is this suggesting that the base shear of a dual system is greater for frames than for shear walls ...?

Thor A. Tandy P.Eng, MIStructE, Struct Eng
Victoria, BC
Canada
vicpeng@telus.net

-----Original Message-----
From: y.hamida [mailto:y.hamida@scs-net.org]
Sent: Thursday, January 10, 2008 11:36 AM
To: seaint@seaint.org
Subject: Dual - System ( frame+wall)

                                                                  Dual - System

        A - code sayes if you have shear walls and ordinary frames (   beams+columns )

                              to resist earthquake
                          You can neglect the ordinary frames and resist the entire base shear by shear walls

                            and designe the ordinary frames just for axial loads.

                           .

       B - M y (ph.d) research sayes;

              if you have  moment resisting frames and walls( reinforce concrete or masonary  )

              to resist earthquake You can neglect the walls and resist the entire base shear

               by  moment resisting frames  and the walls design just for axial loads



             and minimum reinforcement for concrete walls and conect the masonary walls to

                the floor by dowels  .

                        Dr .hamida

RE: Dual - System ( frame+wall)

... surely the stiffer element will take the load up before the frame deforms enough to carry loads ...? Or is this suggesting that the base shear of a dual system is greater for frames than for shear walls ...?

Thor A. Tandy P.Eng, MIStructE, Struct Eng
Victoria, BC
Canada
vicpeng@telus.net

-----Original Message-----
From: y.hamida [mailto:y.hamida@scs-net.org]
Sent: Thursday, January 10, 2008 11:36 AM
To: seaint@seaint.org
Subject: Dual - System ( frame+wall)

Dual - System

A - code sayes if you have shear walls and ordinary frames ( beams+columns )

                              to resist earthquake
                          You can neglect the ordinary frames and resist the entire base shear by shear walls

                            and designe the ordinary frames just for axial loads.

B - M y (ph.d) research sayes;

if you have moment resisting frames and walls( reinforce concrete or masonary )

to resist earthquake You cann neglect the walls and resist the entire base shear

by moment resisting frames and the walls design just for axial loads

and minimum reinforcement for concrete walls and conect the masonary walls to

                the floor by dowels  .

                        Dr .hamida

Dual - System ( frame+wall)

Dual - System

A - code sayes if you have shear walls and ordinary frames ( beams+columns )

B - M y (ph.d) research sayes;

if you have moment resisting frames and walls( reinforce concrete or masonary )

to resist earthquake You cann neglect the walls and resist the entire base shear

by moment resisting frames and the walls design just for axial loads

and minimum reinforcement for concrete walls and conect the masonary walls to

                the floor by dowels  .

                        Dr .hamida

RE: Lighter Loads in IBC vs. UBC

You also need to consider the parapet loads for in in-plane condition,
these values are different than the wall below roof level loads, and
they are much higher in some cases. You will have a parapet load on both
sides of the building, a load on the windward parapet and a load on the
leeward parapet.

Jason

-----Original Message-----
From: Jared Keyser [mailto:jkeyser@LCMF.com]
Sent: Thursday, January 10, 2008 10:18 AM
To: seaint@seaint.org
Subject: RE: Lighter Loads in IBC vs. UBC

Watch out for the Component and Cladding forces. From my experience,
they usually come out much higher.

Jared F. Keyser, P.E.
Anchorage, AK

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*

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RE: UFC 3-340-01 "Design and Analysis of Hardened Structures to Conventional Weapons Effects"

Thomas,
Based on the title of this thread, I assume that you have a copy of the UFC 3-340-01 (AKA TM 5-855-1). You should be looking at Section 10.8.5. The structural steel section is a bit thin based on the amount of research done at the time this publication was developed. Structural steel connections are the last thing you want to fail in a blast design. They should be designed to remain elastic for most all blast resistant conditions.

The rise time of a blast load, the fact that high strength bolts are relatively brittle, and the nature of connection failures make conservativism for connection blast design very prudent. A properly designed window jamb connection for AT/FP is a sight to behold.

There has been some testing done and anecdotal evidence in "normal" structural steel connections subjected to accidental blast loading. You may feel that you are being overly conservative, but when you see the damage in a real incident, you will be comforted by conservativism. Incipient failure is best evidenced in a flexural member not in the connection. Otherwise the flexural member becomes a blast borne missile.

If this (blast engineering) is something in which you are new, I would strongly suggest a peer review by someone seasoned and skilled in blast design. Blast engineering is post-steroidal dynamic design. You may want to check into attending the next SAVIAC conference to learn from the people who really know the topic... (I am not worthy to carry their calculators)

Regards,
Harold Sprague

Date: Thu, 10 Jan 2008 08:09:15 -0700
To: TDavidson@atlasengineering.com
From: gpappas@lanl.gov
Subject: RE: UFC 3-340-01 "Design and Analysis of Hardened Structures to Conventional Weapons Effects"
CC: seaint@seaint.org

On Jan 9, 2008 1:57 PM, Thomas Davidson <TDavidson@atlasengineering.com>
wrote:
> I am new to this blast analysis and was kind of thrown a project. But, I
> was wondering, since this thread came up; if I could get some advice on how
> to analyze bolted connections. I went a conservative route; I got the peak
> overburden pressure and applied it as a wind pressure on the structure
> itself and the "structure" works, the connections however fail miserably.
> Because the load is only going to be applied for a few milliseconds is there
> a justified Dynamic Increase Factor (DIF) for bolts? Or is there some way I
> can justify that the bolts will not fail?
>
> Thanks
>
> *Thomas Davidson*

Put your friends on the big screen with Windows Vista® + Windows Live™. Start now!

RE: Lighter Loads in IBC vs. UBC

Watch out for the Component and Cladding forces. From my experience,
they usually come out much higher.

Jared F. Keyser, P.E.
Anchorage, AK

Re: UFC 3-340-01 "Design and Analysis of Hardened Structures to Conventional Weapons Effects"

Glen,

That is O.K. unless "back in the direction from whence it came" is your boss, employer, payroll check signer. I've been there before. You don't always have a choice.

Joe Grill

----- Original Message -----

From: Glen Pappas

To: Thomas Davidson

Cc: seaint@seaint.org

Sent: Thursday, January 10, 2008 8:09 AM

Subject: RE: UFC 3-340-01 "Design and Analysis of Hardened Structures to Conventional Weapons Effects"

On Jan 9, 2008 1:57 PM, Thomas Davidson <TDavidson@atlasengineering.com>
wrote:
> I am new to this blast analysis and was kind of thrown a project. But, I
> was wondering, since this thread came up; if I could get some advice on how
> to analyze bolted connections. I went a conservative route; I got the peak
> overburden pressure and applied it as a wind pressure on the structure
> itself and the "structure" works, the connections however fail miserably.
> Because the load is only going to be applied for a few milliseconds is there
> a justified Dynamic Increase Factor (DIF) for bolts? Or is there some way I
> can justify that the bolts will not fail?
>
> Thanks
>
> *Thomas Davidson*

Thomas, based on what you've described/asked, I strongly recommend that you 'kind of throw the project back' in the direction from whence it came, & start reading up on blast design (i.e., if think you'd like to actually work on a future one of these thrown your way).

Off the top of my head, Structural Design for Physical Security, State of the Practice, ASCE, 1999, would be a good, readily-available place to start. And the documents referenced therein are all of the 'important' ones.

For recent developments search archives of the following magazines: Modern Steel Construction, Structure, Civil Engineering, The Military Engineer, & Structural Engineer. They've published many good articles on blast design & related topics, some specific to structural steel, since 2003.

Finally, be on the look-out for the soon-to-be published, Blast Protection of Buildings, by ASCE; this will be the 1st national standard on this issue.

Glen

Glen Pappas, Ph.D., PE, SECB, BSCP
Los Alamos National Laboratory
ES-DE
Technical Staff Member
Phone: 505-665-1221
Fax: 505-665-4728
MS M791
TA-00-0786
gpappas@lanl.gov

RE: UFC 3-340-01 "Design and Analysis of Hardened Structures to Conventional Weapons Effects"

On Jan 9, 2008 1:57 PM, Thomas Davidson <TDavidson@atlasengineering.com>
wrote:
> I am new to this blast analysis and was kind of thrown a project. But, I
> was wondering, since this thread came up; if I could get some advice on how
> to analyze bolted connections. I went a conservative route; I got the peak
> overburden pressure and applied it as a wind pressure on the structure
> itself and the "structure" works, the connections however fail miserably.
> Because the load is only going to be applied for a few milliseconds is there
> a justified Dynamic Increase Factor (DIF) for bolts? Or is there some way I
> can justify that the bolts will not fail?
>
> Thanks
>
> *Thomas Davidson*

Thomas, based on what you've described/asked, I strongly recommend that you 'kind of throw the project back' in the direction from whence it came, & start reading up on blast design (i.e., if think you'd like to actually work on a future one of these thrown your way).

Off the top of my head, Structural Design for Physical Security, State of the Practice, ASCE, 1999, would be a good, readily-available place to start. And the documents referenced therein are all of the 'important' ones.

For recent developments search archives of the following magazines: Modern Steel Construction, Structure, Civil Engineering, The Military Engineer, & Structural Engineer. They've published many good articles on blast design & related topics, some specific to structural steel, since 2003.

Finally, be on the look-out for the soon-to-be published, Blast Protection of Buildings, by ASCE; this will be the 1st national standard on this issue.

Glen

Glen Pappas, Ph.D., PE, SECB, BSCP
Los Alamos National Laboratory
ES-DE
Technical Staff Member
Phone: 505-665-1221
Fax: 505-665-4728
MS M791
TA-00-0786
gpappas@lanl.gov

Wednesday, January 9, 2008

Re: Out of Line Building Official?

I disagree.

We are not allowed to renew our licenses unless it's within 2 months of the expiration date. So the plans sit in some box for two months (or as happens way more than it should, it gets lost) till the plan checker gets all the other end of the year submittals off their desk and now the client needs to pay for more prints, run around and get new signatures, and the engineer needs to drop everything because the city is busy?

And this stupid policy that building departments have that entire sets need to be resubmitted when one discipline makes a minor change is just wasteful.

-gm

On Jan 9, 2008 7:11 PM, sscholl2@juno.com <sscholl2@juno.com> wrote:

This request of the Bldg. Official is reasonable. This is easily fixed. My problems are about 100 times more significant.

Stan Scholl, P.E.
Laguna Beach, CA
_____________________________________________________________
Click to shop for certified diamonds and engagement rings.
http://thirdpartyoffers.juno.com/TGL2111/fc/Ioyw6iifzDNhN4AyA9qw2HJyONJHShZUXPJzeFIARFNJLe0L4peLaL/

******* ****** ******* ******** ******* ******* ******* ***
* Read list FAQ at: http://www.seaint.org/list_FAQ.asp
*
* This email was sent to you via Structural Engineers
* Association of Southern California (SEAOSC) server. To
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*
* http://www.seaint.org/sealist1.asp
*
* Questions to seaint-ad@seaint.org. Remember, any email you
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--
-gm

Re: Out of Line Building Official?

This request of the Bldg. Official is reasonable. This is easily fixed. My problems are about 100 times more significant.

Stan Scholl, P.E.
Laguna Beach, CA
_____________________________________________________________
Click to shop for certified diamonds and engagement rings.
http://thirdpartyoffers.juno.com/TGL2111/fc/Ioyw6iifzDNhN4AyA9qw2HJyONJHShZUXPJzeFIARFNJLe0L4peLaL/

Re: Lighter Loads in IBC vs. UBC and Enercalc 6.0

When did you receive it?

I haven't received my copy yet.

S.Macie, P.E., SLO, CA

In a message dated 1/9/2008 4:36:09 P.M. Pacific Standard Time, Pinyonengineering@hughes.net writes:

Hi,

with that transition to the 2007 California Builidng Code - I just got Enercalc 6.0 and looked at the wind module and it references the 2003 IBC and some provisions that are not in the 2006 (or ASCE 7-05) seems like the design wind pressure part of the module won't calculate

Has anyone used this program yet for use with the 2006 IBC code?

I am also confused in the seismic module in that I have a redundancy factor of 1.3 and it seems like I need to actually math any use a pencil :-) to put this factor in the calculations. I know this a new version and a new code but I seem lost. I am also doing this all by hand to verify what is supposed to happen. Is there anyone to set me straight??

Tim Rudolph

Pinyon Engineering

Bishop CA

Start the year off right. Easy ways to stay in shape in the new year.

Re: Lighter Loads in IBC vs. UBC and Enercalc 6.0

In a message dated 1/9/2008 4:22:31 PM Pacific Standard Time, Pinyonengineering@hughes.net writes:

Hi,

with that transition to the 2007 California Buildng Code - I just got Enercalc 6.0 and looked at the wind module and it references the 2003 IBC and some provisions that are not in the 2006 (or ASCE 7-05) seems like the design wind pressure part of the module won't calculate

Has anyone used this program yet for use with the 2006 IBC code?

I am also confused in the seismic module in that I have a redundancy factor of 1.3 and it seems like I need to actually math any use a pencil :-) to put this factor in the calculations. I know this a new version and a new code but I seem lost. I am also doing this all by hand to verify what is supposed to happen. Is there anyone to set me straight??

Tim Rudolph

Pinyon Engineering

Bishop CA

Tim

There should be a yes or no question in regard to the following if the program does not calculate this.

Looking at each building direction north=south and east-west If you remove any lateral resisting element (shear wall, frame..) does the removal result in a story strength reduction of more than 33% of the over all story strength.

If the answer is no you can use a rho of 1.0. You have to evaluate both directions.

Now if you have multistory building you check this at all levels that have greater than 35% of the base shear load at that level. If no level has greater than 35% base shear then you can use a rho of 1.0.

Also this check is for SDC of D thru F ...SDC of C and below rho = 0.

Joe Venuti
Johnson & Nielsen Associates
Palm Springs, CA

Start the year off right. Easy ways to stay in shape in the new year.

RE: Lighter Loads in IBC vs. UBC and Enercalc 6.0

I'm still waiting for my email notification to get the update....

YI YANG, S.E.

From: Pinyon Engineering [mailto:Pinyonengineering@hughes.net]
Sent: Wednesday, January 09, 2008 3:54 PM
To: seaint@seaint.org
Subject: RE: Lighter Loads in IBC vs. UBC and Enercalc 6.0

Hi,

with that transition to the 2007 California Builidng Code - I just got Enercalc 6.0 and looked at the wind module and it references the 2003 IBC and some provisions that are not in the 2006 (or ASCE 7-05) seems like the design wind pressure part of the module won't calculate

Has anyone used this program yet for use with the 2006 IBC code?

I am also confused in the seismic module in that I have a redundancy factor of 1.3 and it seems like I need to actually math any use a pencil :-) to put this factor in the calculations. I know this a new version and a new code but I seem lost. I am also doing this all by hand to verify what is supposed to happen. Is there anyone to set me straight??

Tim Rudolph

Pinyon Engineering

Bishop CA

RE: Lighter Loads in IBC vs. UBC and Enercalc 6.0

Hi,

Has anyone used this program yet for use with the 2006 IBC code?

Tim Rudolph

Pinyon Engineering

Bishop CA

Re: UFC 3-340-01 "Design and Analysis of Hardened Structures to Conventional Weapons Effects"

Are you looking at bolted connections of steel? The 9th edition ASD steel manual allowed a 200% load duration factor for impact loads. Don't know off-hand if ACI has a similar "strength" factor... good luck.

On Jan 9, 2008 1:57 PM, Thomas Davidson <TDavidson@atlasengineering.com> wrote:

I am new to this blast analysis and was kind of thrown a project. But, I was wondering, since this thread came up; if I could get some advice on how to analyze bolted connections. I went a conservative route; I got the peak overburden pressure and applied it as a wind pressure on the structure itself and the "structure" works, the connections however fail miserably. Because the load is only going to be applied for a few milliseconds is there a justified Dynamic Increase Factor (DIF) for bolts? Or is there some way I can justify that the bolts will not fail?

Thanks

Thomas Davidson

From: Padmanabhan Rajendran [mailto:rakamaka@yahoo.com]
Sent: Wednesday, January 09, 2008 12:12 PM
To: seaint@seaint.org
Subject: Re: UFC 3-340-01 "Design and Analysis of Hardened Structures to Conventional Weapons Effects"

In 2006, I attended a seminar on the design of structures to resist blast loads. The instructor told the participants that a few of the UFC/UASCE manuals which were available for download until 2002 or 2003 were removed from the list of manuals that could be downloaded with "no questions asked". However, he said, that anyone interested in a restricted manual could seek permission from UFC/UASCE for the download. I believe the identity of the individual requesting the download requires to be documented, considering the nature of information contained in the manuals.

The following link has a place for you to register. Once you register, you may be able to access the document.
https://pdc.usace.army.mil/forums/ufc/3-340-01

Rajendran

----- Original Message ----
From: Bill Polhemus <bill@polhemus.cc>
To: seaint@seaint.org
Sent: Tuesday, January 8, 2008 1:53:07 PM
Subject: UFC 3-340-01 "Design and Analysis of Hardened Structures to Conventional Weapons Effects"

I am doing a blast analysis of a structure located in a refinery, using
SBEDS which is an Excel-based software solution maintained by the USACE.
In a couple of places in the documentation, they references the manual
as stated in the SUBJECT line of my message.

However, this publication isn't available online, unless you're a U.S.
Gov't contractor actively working on government jobs. This is
problematic, since much of the methodology in SBEDS is contained therein.

For example in one place they reference "the ultimate moment capacity of
a reinforced concrete beam-column ... as defined by equations in Chapter
10 of UFC 3-340-01." I would seriously love to examine those equations
but cannot as far as I know, because, again, I'm not privy to this
information.

Has anyone run up against this or something like it, and how do you
LEGALLY get around this restriction? Strange that a program in the
public domain should be based on methodology that is literally a
"federal secret."

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David Topete, SE

Re: Lighter Loads in IBC vs. UBC

Chris-
Paul is correct in you being not incorrect. Our office has done comparisons of wind and seismic and have come to similar conclusions. We've found that wind is more cumbersome to arrive at a value that's about 25% less than the UBC value derived from a single equation...

On Jan 9, 2008 12:34 PM, Paul Feather <PFeather@se-solutions.net> wrote:

You are not incorrect.

The UBC wind was originally derived from ASCE 7 with conservative
assumptions, and therefore when deriving actual ASCE 7 values they will
typically be slightly lower (regardless of the three extra days to
derive the value).

Seismic is also going to be lower in many instances, but not all. Under
UBC all structures in a "zone" were equally penalized. Under the IBc
provisions the seismic requirements are much more site specific, and in
most cases will be lower until you reach the type of site that set the
former threshold for the zone.

Paul Feather PE, SE
pfeather@SE-Solutions.net
www.SE-Solutions.net

-----Original Message-----
From: chris.slater@gmail.com [mailto:chris.slater@gmail.com] On Behalf
Of Chris Slater
Sent: Wednesday, January 09, 2008 12:29 PM
To: seaint@seaint.org
Subject: Lighter Loads in IBC vs. UBC

So I think I need a sanity check. Last time I asked a question here I
got blasted at first and then wound up getting some very useful
information. So I suppose I'll give it another shot!

We're working on making the transition from 1997 UBC to 2006 IBC.
While the wind loads are more complicated, I'm finding that as long as
topography isn't a factor, even the highest load (that of the end
conditions) is the same or lower than what we were using in the UBC.
So we can simplify and use the "A" load across the whole projected
area, and still generate Wind loads that are lower than what we were
designing for before.

Similarly, when I go through the USGS software and get values for S1
and Ss, then put that through the equations in ASCE7-05, I come up
with a Base Shear equation that is a fair bit lower. Where a one
story home in the Bay Area would have previously had a V of say 0.14W,
the new code puts it at 0.108W.

I can send more details of how I'm coming up with these figures, but I
just wanted to start by finding out if other folks are seeing the same
thing.

Thanks in advance.

Chris Slater, PE

******* ****** ******* ******** ******* ******* ******* ***
* Read list FAQ at: http://www.seaint.org/list_FAQ.asp
*
* This email was sent to you via Structural Engineers
* Association of Southern California (SEAOSC) server. To
* subscribe (no fee) or UnSubscribe, please go to:
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--
David Topete, SE

RE: UFC 3-340-01 "Design and Analysis of Hardened Structures to Conventional Weapons Effects"

I am new to this blast analysis and was kind of thrown a project. But, I was wondering, since this thread came up; if I could get some advice on how to analyze bolted connections. I went a conservative route; I got the peak overburden pressure and applied it as a wind pressure on the structure itself and the "structure" works, the connections however fail miserably. Because the load is only going to be applied for a few milliseconds is there a justified Dynamic Increase Factor (DIF) for bolts? Or is there some way I can justify that the bolts will not fail?

Thanks

Thomas Davidson

From: Padmanabhan Rajendran [mailto:rakamaka@yahoo.com]
Sent: Wednesday, January 09, 2008 12:12 PM
To: seaint@seaint.org
Subject: Re: UFC 3-340-01 "Design and Analysis of Hardened Structures to Conventional Weapons Effects"

In 2006, I attended a seminar on the design of structures to resist blast loads. The instructor told the participants that a few of the UFC/UASCE manuals which were available for download until 2002 or 2003 were removed from the list of manuals that could be downloaded with "no questions asked". However, he said, that anyone interested in a restricted manual could seek permission from UFC/UASCE for the download. I believe the identity of the individual requesting the download requires to be documented, considering the nature of information contained in the manuals.

The following link has a place for you to register. Once you register, you may be able to access the document.
https://pdc.usace.army.mil/forums/ufc/3-340-01

Rajendran

----- Original Message ----
From: Bill Polhemus <bill@polhemus.cc>
To: seaint@seaint.org
Sent: Tuesday, January 8, 2008 1:53:07 PM
Subject: UFC 3-340-01 "Design and Analysis of Hardened Structures to Conventional Weapons Effects"

I am doing a blast analysis of a structure located in a refinery, using
SBEDS which is an Excel-based software solution maintained by the USACE.
In a couple of places in the documentation, they references the manual
as stated in the SUBJECT line of my message.

However, this publication isn't available online, unless you're a U.S.
Gov't contractor actively working on government jobs. This is
problematic, since much of the methodology in SBEDS is contained therein.

For example in one place they reference "the ultimate moment capacity of
a reinforced concrete beam-column ... as defined by equations in Chapter
10 of UFC 3-340-01." I would seriously love to examine those equations
but cannot as far as I know, because, again, I'm not privy to this
information.

Has anyone run up against this or something like it, and how do you
LEGALLY get around this restriction? Strange that a program in the
public domain should be based on methodology that is literally a
"federal secret."

******* ****** ******* ******** ******* ******* ******* ***
* Read list FAQ at: http://www.seaint.org/list_FAQ.asp
*
* This email was sent to you via Structural Engineers
* Association of Southern California (SEAOSC) server. To
* subscribe (no fee) or UnSubscribe, please go to:
*
* http://www.seaint.org/sealist1.asp
*
* Questions to seaint-ad@seaint.org. Remember, any email you
* send to the list is public domain and may be re-posted
* without your permission. Make sure you visit our web
* site at: http://www.seaint.org
******* ****** ****** ****** ******* ****** ****** ********

Be a better friend, newshound, and know-it-all with Yahoo! Mobile. Try it now.

RE: Lighter Loads in IBC vs. UBC

You are not incorrect.

The UBC wind was originally derived from ASCE 7 with conservative
assumptions, and therefore when deriving actual ASCE 7 values they will
typically be slightly lower (regardless of the three extra days to
derive the value).

Seismic is also going to be lower in many instances, but not all. Under
UBC all structures in a "zone" were equally penalized. Under the IBc
provisions the seismic requirements are much more site specific, and in
most cases will be lower until you reach the type of site that set the
former threshold for the zone.

Paul Feather PE, SE
pfeather@SE-Solutions.net
www.SE-Solutions.net

-----Original Message-----
From: chris.slater@gmail.com [mailto:chris.slater@gmail.com] On Behalf
Of Chris Slater
Sent: Wednesday, January 09, 2008 12:29 PM
To: seaint@seaint.org
Subject: Lighter Loads in IBC vs. UBC

So I think I need a sanity check. Last time I asked a question here I
got blasted at first and then wound up getting some very useful
information. So I suppose I'll give it another shot!

We're working on making the transition from 1997 UBC to 2006 IBC.
While the wind loads are more complicated, I'm finding that as long as
topography isn't a factor, even the highest load (that of the end
conditions) is the same or lower than what we were using in the UBC.
So we can simplify and use the "A" load across the whole projected
area, and still generate Wind loads that are lower than what we were
designing for before.

Similarly, when I go through the USGS software and get values for S1
and Ss, then put that through the equations in ASCE7-05, I come up
with a Base Shear equation that is a fair bit lower. Where a one
story home in the Bay Area would have previously had a V of say 0.14W,
the new code puts it at 0.108W.

I can send more details of how I'm coming up with these figures, but I
just wanted to start by finding out if other folks are seeing the same
thing.

Thanks in advance.

Chris Slater, PE

******* ****** ****** ****** ******* ****** ****** ********