Structural Engineering: 2008-06-08

Saturday, June 14, 2008

Re: Truck Impact Load on the Wall

Around a year ago, IIRC, ENR reported that an errant truck had knocked a large concrete supporting column out from below an overpass on a Texas freeway. No protective measures whatsoever! Surprised the heck out of me; I would have thought that such critical members would be either protected or have the strength to survive such predictible impacts. I wonder if the AASHTO requirements you mention have "always" been in effect.

Ralph Hueston Kratz, S.E.
Richmond CA

In a message dated 6/14/08 2:21:09 PM, jake.watson1@gmail.com writes:

Sorry, I'm a little late and hope this is still useful. The AASHTO bridge code as several (6 I believe) impact levels based on the geometry of the road and speed of the vehicle. AASHTO's impact forces can be quite severe and can exceed 100k in some of the worst cases. I suggest you dig into those provisions.

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Re: Truck Impact Load on the Wall

Sorry, I'm a little late and hope this is still useful. The AASHTO bridge code as several (6 I believe) impact levels based on the geometry of the road and speed of the vehicle. AASHTO's impact forces can be quite severe and can exceed 100k in some of the worst cases. I suggest you dig into those provisions.

Jake Watson, S.E.
Salt Lake City, UT

On Thu, Jun 5, 2008 at 11:06 AM, Wontae Kim <kimwontae@email.com> wrote:

Hi,

IBC 1607.7.3 refers only to passenger car impact load conditions.

What are 'reasonable' impact load conditions for precast concrete parking garage which accommodates HS20 trucks?

Thanks!

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Re: Seismic loads using ASCE 7

In a message dated 6/6/2008 9:41:10 AM Pacific Daylight Time, davea@laneengineers.com writes:

One of the things that we are sort of struggling with here (old "zone 3" country) is that we are finding seismic loads via ASCE 7 wind up being less (sometimes uncomfortably) than what we used to calculate using the 1997 UBC. As a result, we've sort of adopted our own internal "company minimum seismic coefficient" that we are comfortable with. Are most people discovering the same thing and what are your thoughts?

You are not the only ones to find out lower values with the new code (smile).

Antonio S. Luisoni

Consulting SE

Granada Hills, CA

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Friday, June 13, 2008

Re: affecting your bottom line?

In answer to Dennis' question: there is no slowdown in my engineering work coming in. Most are additions to houses on the west side of LA and owner's representative for tenant improvements in multi-story office buildings.

Stan Scholl, P.E.
Laguna Beach, CA

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RE: moment frame achorage

Thanks Mark, I appreciate that. What you say makes sense.
I would then think that the grade beam could be modeled a beam, with the
moment of the column resolved in to two point loads. The load from the
compression flange would be one support. The tension in the anchor bolts
would be a point load very close to the support. The other support would be
at the other end of the grade beam. Stirrups could be provided to resist
the shear induced in the grade beam by the point load due to the anchor
bolts (neglecting the idea that it is within the depth of the beam to the
support).
Does that sound reasonable?

Thanks

-----Original Message-----
From: Mark Gilligan [mailto:m_k_gilligan@yahoo.com]
Sent: Friday, June 13, 2008 9:24 AM
To: seaint@seaint.org
Subject: RE: moment frame achorage

The secret to dealing with Appendix D is to realize that the provisions were
developed to deal with anchors in essentially unreinforced concrete. This
is not the case with a grade beam.
Think of the connection as a D region that is resolved using a system of
struts and ties. The tension in the anchor rods is a tie that is resolved
by one or more struts that in turn engage the stirrups in the grade beam.
I believe that the ACI 318-08 implicitly recognizes this approach.
Regarding the plates at the bottom of the anchor rod. They should be sized
so that relatively little force is transfered by bending in the plates. The
forces want to be resolved close to the rod.
Mark Gilligan

******* ****** ****** ****** ******* ****** ****** ********

Re: How are Subscription Costs for Engineering Software and CAD affecting your bottom line?

OK Dennis, I'll take this one on. I'm fairly passionate about
software and your message has been rattling around in my head since I
read it yesterday.

While those prices don't sound horrible to me, I can certainly see
where slow times like these put pressure on to reduce costs wherever
possible. We don't use TEDDS, but it seems to me like $150 software
could pay for itself fairly quickly, assuming you have enough work to
keep you busy. Your time is worth a lot. At $350, it has a lot more
work to do to be worth it's price tag.

Of course, if the work isn't coming in, then you can "afford" to spend
more time, and less money to get each job done.

Other than CAD and Enercalc, all of the software we use is custom
scripts that I've written to automate our calculations. Those use
Excel (which is a one time purchase and fairly necessary, though
OpenOffice would work as well) and Perl, which is free. The nice
thing is that if we want to add an additional calc, or change the way
we do things, all I have to do is change my script.

I've looked into competitors to AutoCAD and while there's a few out
there, it seems like they all have a long way to go. Some run on
Linux, and after just spending two days de-virusing a couple of our
workstations, it's tempting to make the switch, but the lack of
AutoCAD pretty well shuts that door.

Not sure if I've answered any of your questions, or just rambled. My
$0.02 at any rate.

Chris

Chris Slater, PE
JDWylie Engineering, Inc
(209) 577-2339
Home Office: (530) 268-1440
Cell Phone: (916) 303-0889

On Thu, Jun 12, 2008 at 11:44 AM, Dennis Wish <dennis.wish@verizon.net> wrote:
> I recently decided I had to end my subscription to TEDDS Maintenance &
> Update (MU) agreement. It appears that they raised the price from the last
> two years for their MU from $150.00 per year for TEDDS during the 2006 year
> to approximately $250.00 per year for Release 10 & 11 and now $350.00 per
> year for the M&U on the current Release version that was recently sent out –
> TEDDS v. 11. As with AutoCad and other software developers, they will
> reinstate your MU if you renew with the next three years based on the fee
> times the number of years expired. So in three years, you can reinstate to
> the most current version of TEDDS for 3x$350.00 (if the annual fee does not
> change in that time) or $1,050.00. This is currently approximately the cost
> of a new seat.
>
>
>
> As a small business owner I have had to make a decision. Business is down,
> the bottom line is down and the Codification of the 2006 IBC and ASCE 7-05
> this year has he hard. It is not an issue of competition since the majority
> of us in my area are hurting for new work of any substance. I have decided
> to let AutoCad go for this year, but will probably try to reinstate the year
> since the cost before Autodesk orphans the software is similar to TEDDS – in
> my case for AutoCad Architecture 2008; $595.00 + tax times the number of
> years delinquent. There is no comparison, I could live without TEDDS and may
> go back to a prior version of MathCAD™ which would be more competitive and
> do the work that TEDDS does without the library that I have mixed feelings
> over. AutoCad is more important in the operation of my business than TEDDS
> but with these companies needs to keep cash flow moving through the company
> for development they will charge increasing annual M&U fees that are
> cumulative is simply, in my opinion, a bad marketing decision for small
> office engineers.
>
>
>
> Does this strike you as a potential problem for small offices who are harder
> hit on their bottom line with the increased cost of tools and references?
> Would you believe that if small offices are driven out of business that the
> large firms in the US will be competitive with base offices in the US who
> outsource their design and analysis work to China and India or other
> countries that have a much lower labor rate?
>
>
>
> I'm interested in whether or not many of you are starting to see a hit from
> the current economic situation that is causing you to tighten your belts and
> give up tools that you once felt were your competitive edge. In fact, is
> this idea of software becoming the reviving "Competitive edge" a resurging
> prospect?
>
>
>
> Dennis S. Wish, PE
>
>
>
> Dennis S. Wish, PE
>
> California Professional Engineer
>
> Structural Engineering Consultant
>
> La Quinta, CA 92253
>
> 760.564.0884 (Phone, Fax and Answering Machine)
>
> dennis.wish@verizon.net
>
> http://structuralist.wordpress.com
>
> http://www.structuralist.net
>
>

RE: moment frame achorage

The secret to dealing with Appendix D is to realize that the provisions were developed to deal with anchors in essentially unreinforced concrete. This is not the case with a grade beam.
Think of the connection as a D region that is resolved using a system of struts and ties. The tension in the anchor rods is a tie that is resolved by one or more struts that in turn engage the stirrups in the grade beam.
I believe that the ACI 318-08 implicitly recognizes this approach.
Regarding the plates at the bottom of the anchor rod. They should be sized so that relatively little force is transfered by bending in the plates. The forces want to be resolved close to the rod.
Mark Gilligan

Thursday, June 12, 2008

How are Subscription Costs for Engineering Software and CAD affecting your bottom line?

I recently decided I had to end my subscription to TEDDS Maintenance & Update (MU) agreement. It appears that they raised the price from the last two years for their MU from $150.00 per year for TEDDS during the 2006 year to approximately $250.00 per year for Release 10 & 11 and now $350.00 per year for the M&U on the current Release version that was recently sent out – TEDDS v. 11. As with AutoCad and other software developers, they will reinstate your MU if you renew with the next three years based on the fee times the number of years expired. So in three years, you can reinstate to the most current version of TEDDS for 3x$350.00 (if the annual fee does not change in that time) or $1,050.00. This is currently approximately the cost of a new seat.

As a small business owner I have had to make a decision. Business is down, the bottom line is down and the Codification of the 2006 IBC and ASCE 7-05 this year has he hard. It is not an issue of competition since the majority of us in my area are hurting for new work of any substance. I have decided to let AutoCad go for this year, but will probably try to reinstate the year since the cost before Autodesk orphans the software is similar to TEDDS – in my case for AutoCad Architecture 2008; $595.00 + tax times the number of years delinquent. There is no comparison, I could live without TEDDS and may go back to a prior version of MathCAD™ which would be more competitive and do the work that TEDDS does without the library that I have mixed feelings over. AutoCad is more important in the operation of my business than TEDDS but with these companies needs to keep cash flow moving through the company for development they will charge increasing annual M&U fees that are cumulative is simply, in my opinion, a bad marketing decision for small office engineers.

Does this strike you as a potential problem for small offices who are harder hit on their bottom line with the increased cost of tools and references? Would you believe that if small offices are driven out of business that the large firms in the US will be competitive with base offices in the US who outsource their design and analysis work to China and India or other countries that have a much lower labor rate?

I’m interested in whether or not many of you are starting to see a hit from the current economic situation that is causing you to tighten your belts and give up tools that you once felt were your competitive edge. In fact, is this idea of software becoming the reviving “Competitive edge” a resurging prospect?

Dennis S. Wish, PE

California Professional Engineer

Structural Engineering Consultant

La Quinta, CA 92253

760.564.0884 (Phone, Fax and Answering Machine)

dennis.wish@verizon.net

http://structuralist.wordpress.com

http://www.structuralist.net

Anchor Bolt Shear Capacity in CMU

Have any of you run through the calculations for the shear capacity of an anchor bolt at the top of a CMU wall using section 2.1.4.2.3 of ACI 530-5. This is a situation where for transverse loads, the side distance is less than 12 bolt diameters, so you have to discount the shear capacity of the masonry.

I was checking the capacity of an anchor bolt to plate connection, but increasing the bolt diameter actually reduced the capacity. There is an interpolation requirement based on side distance that actually ends up penalizing you for increasing the bolt size. This occurs because the allowable capacity is taken as the maximum capacity times the factor (side distance minus 1”)/(12 bolt diameters minus 1”) and this factor decreases with bolt diameter faster than the maximum capacity increases with bolt diameter. Maybe I’m missing something, but why should you lose capacity when the bolt size goes up. Can I pretend it’s a smaller bolt? Can I use two bolts offset each side of the wall centerline and base the side distance on the far face?

Jim Lutz, P.E., S.E.

720 3rd Avenue, Suite 1200

Seattle, WA 98104-1820

206 505 3400 Ext 126

206 505 3406 (Fax)

www.bhcconsultants.com

Re: 1985 pre-manufactured building

ron buchko wrote:
> What specific "general notes" lead you to believe that these existing
> roof rurlins will NOT comply?
>
> --- On *Wed, 6/11/08, Drew Morris /<dmorris@bbfm.com>/* wrote:
>
> From: Drew Morris <dmorris@bbfm.com>
> Subject: 1985 pre-manufactured building
> To: "SEAINT" <seaint@seaint.org>
> Date: Wednesday, June 11, 2008, 7:10 PM
>
>
The General Notes give the design snow load of 38 psf. I can get the
purlins to only work at Fy=50 ksi, but not at 33 ksi.
New buildings in the area are designed for a minimum of 60 psf, so we
are evaluating the options for upgrading the building.
The frames are of more concern, I know the purlins need to be infilled.

******* ****** ****** ****** ******* ****** ****** ********

Re: Wind Load on Deep Snow Profile

The code load case is 75% wind with 75% snow. While, intuitively, the high winds should blow the snow off the roof depending on the fluffiness of the snow, the code doesn't allow you reduce the snow load less than 75%.

An old quote from a regulatory agency "Don't confuse reason with requirements."

Paul.

On 6/11/08, Jake Watson <jake.watson1@gmail.com> wrote:

One other thought, as the wind presure increases, will it blow the
snow off the roof?  If the drift is parallel to the wind load, then
there is not an increase in area.  This is the typical case considered
in the code.  If the wind blows the other way, the drift is reduced.
The wind won't load the drift zone in the first place.  Any drift
caused by wind blowing the other direction will be moved again by the
new wind direction.

This really needs a picture.

Jake Watson SE
Salt lake city, UT

On 6/10/08, Paul Blomberg <paul.blomberg@gmail.com> wrote:
> I'm putting together design criteria and am curious how you handle / take
> into consideration the added projected wind surface area due to built up
> snow on the roof of a building?
>
> I have a new metal building going into northwestern Montana and the ground
> snow load is over 120 psf.  That is a relatively deep pile of snow.  When
> you put together your load case that combines snow plus wind, do you
> increase the projected profile of the building to reflect the height of the
> snow on the roof and the associated wind load?
>
> I normally include this additional snow profile to calc the wind loads but
> this my first heavy snow load situation where the increased load is
> significant.
>
> Paul.
> Phoenix, AZ
>

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Re: 1985 pre-manufactured building

What specific "general notes" lead you to believe that these existing roof rurlins will NOT comply?

--- On Wed, 6/11/08, Drew Morris <dmorris@bbfm.com> wrote:

From: Drew Morris <dmorris@bbfm.com>
Subject: 1985 pre-manufactured building
To: "SEAINT" <seaint@seaint.org>
Date: Wednesday, June 11, 2008, 7:10 PM

I am evaluating an existing pre-manufactured building that was designed  in 1985 under the 1982 UBC.  The roof purlins are 12 gage.  Is it safe  to assume that these are Fy = 50 ksi?  At 33 ksi, they don't meet the  loads in the General Notes.  Also, would the columns and frames be A36?   A572 Gr 50 was available then, but I am not sure if that material was  used by the manufacturers.  Would it have been cheaper to use thicker  and wider plates of A36 versus some material with Fy = 50 ksi.  ******* ****** ******* ******** ******* ******* ******* *** *   Read list FAQ at: http://www.seaint.org/list_FAQ.asp *  *   This email was sent to you via Structural Engineers  *   Association of Southern California (SEAOSC) server. To  *   subscribe (no fee) or UnSubscribe, please go to: * *   http://www.seaint.org/sealist1.asp * *   Questions to seaint-ad@seaint.org. Remember, any email you  *   send to the list is public domain and may be re-posted  *   without your permission. Make sure you visit our web  *   site at: http://www.seaint.org  ******* ****** ****** ****** ******* ****** ****** ********

moment frame achorage

Sirs:

I am designing anchorage for a fixed moment frame base.
"Years ago" I was taught to place the bottom of the anchor bolts at the
bottom of a grade beam. The footing was deepened in this area to provide
coverage. Plate washers were placed on each bolt. The plate washers were
analyzed like a base plate: big enough to provide adequate bearing, thick
enough for flexure and placed deep enough to accommodate punching shear and
one way bending shear in the grade beam.

Now there is Appendix D and its coverage requirements. With its assumed
failure plane, it seems that it would take a 9 foot square x 3 foot thick
piece of concrete just to fully anchor a single 3 foot deep anchor bolt.
There are reductions for being "near the edge", but they are rather severe.
Is this true? (this analysis does ignore the down load from the compression
side of the column). How do others determine this anchorage? Is there a
good recent example?

Thanks again, in advance, for any of your thoughts.

Joseph Eribarne
661-392-1124
jeribarne@bak.rr.com

******* ****** ****** ****** ******* ****** ****** ********

RE: 1 or 2 piles under a cap

"stone clad pyramidal piers" - I have never used something like this.

Helical pier installation, in my experience, is actually quite easy, and
even more so in a strictly vertical installation, and it is very accurate.
If the pier is 3-4" off center, then it's time to get another pier
installer. I've used helical piers in 8" wide grade beams, cast within the
grade beam using the "new construction load transfer bracket" (a chunk of
tube steel with a couple #5 rebar sticking up, kinda like rabbit ears).

This bracket was tested using a 3,000 psi concrete (cylandars testes at 7,
14, and 28 broke at 3250, 4090, and 4780 respecitvely) to a total load of
106K. The bracket was HSS2x2x1/4 by 6" with a top plate of 2" by 4" by 1/2"
and two (2) #5 rebar by 12" long welded to the top plate and HSS. This was
cast in a 10" sonotube and sent to compression testing. The unreinforced
concrete failed under compression testing, and it sheared from all
imbedments. In short, the concrete failed, the bracket did not.

If you build a design capacity 50K pier (100K ulitimate), and it is strictly
vertical load, then you could do a 12" by 18" to 24" sonotube pier cap at
the location of the load to react it specifically. Use other, smaller,
piers to cover holding the weight of the precast concrete deck and any
lateral loads you may have.

I've used piers for 25K and 50K capacity for foundation remediation and new
foundation construction, as well as isolated conditions that you are
experiencing. The installation is quite accurate, and again, if you have a
contractor that misses the target by 3" to 4", they shouldn't be installing
piers to begin with. If they don't meet location requirements, have them
take the piers out and try again (at the contractor's dime of course).
After a couple of zinged piers, he's going to get the idea that he is either
going to have to do it right, or go broke doing it "they way they always
have." *laugh*

Does this help at all?

Dave Maynard
Gillette, WY

-----Original Message-----
From: Jordan Truesdell, PE [mailto:seaint2@truesdellengineering.com]
Sent: Wednesday, June 11, 2008 7:31 PM
To: seaint@seaint.org
Subject: Re: 1 or 2 piles under a cap

These are isolated footings supporting a precast exterior deck. It's a
very simple condition, actually - much like a residential deck, but with
precast plank and foolishly large piers. The piers may drop to 16"
sonotube when the client gets the estimate on the stone-clad, pyramidal
piers. I suppose I could just encase the top of the helical pier in the
concrete pier and forgo the separate pile cap, but I'm concerned about
the potential for moment in the helical pier shaft if the installation
isn't perfectly centered under the post above. 3-4" off with a 40+kip
load would likely exceed the capacity of the shaft. Most of the contact
I've had with these types of piles is either as shoring of existing
buildings with light loads (maybe 10k), or using multiples under caps
for capacity reasons.

Jordan

David Maynard wrote:
> John Pack, PE, of IMR, Inc. out of Denver, CO, has done extensive studies
> with AB Chance Helical Piers. If you do an installation torque of 10,000
> ft. lbs, you could get an Ultimate Capacity of the pier to be 100 kips.
> With a factor of safety of 2.0, that's a 50 kips design capacity. If they
> have the materials to build a 50 kip (design capacity, 100 kip ultimate
> capacity) helical pier (heavier shaft, several helix flights, etc., then
you
> can use a local helical pier for this particular load and isolate it from
> the rest of the foundation.
>
> Otherwise, with the spread that you have, you would have to design the
> foundation as a concrete beam between these two piers. I might not have a
> clear understanding of your problem. If you had a diagram to share, I
might
> have more information.
>
> Dave Maynard
> Gillette, WY
>
>
> -----Original Message-----
> From: Jordan Truesdell, PE [mailto:seaint2@truesdellengineering.com]
> Sent: Wednesday, June 11, 2008 9:28 AM
> To: seaint@seaint.org
> Subject: 1 or 2 piles under a cap
>
> Does anyone have a rule of thumb for using less than 3 piles (helicals,
> in my case) under a pile cap? I've got a condition where I have some
> lightly loaded pile caps - about 45k - but their supporting pyramidal
> piers which have a 40" square base. The tops of the piers will support
> a beam line and concrete plank, so there's no lateral loads. The piers
> are 16-20' apart and are in a (roughly) straight line, so I can't
> efficiently combine multiple piers under a single footing.
>
>

******* ****** ****** ****** ******* ****** ****** ********

Argintina Wind Speed - For engineers familiar with CIRSOC 102 and ASCE 7-05

Hi,

The followings are from copy from CIRSOC 102.

Speed equal to the average speed (peak gust) intervals of about 3 seconds, on exhibition open to a normal height of 10 m reference that has a recurrence period of one year according to the type and destination of this .

My question is if this wind speed is obtained based on the same condition as ASCE 7 - 05: Values are nominal design 3-second gust wind speeds in m/s at 10m above ground for exposure C catefory.

If not, how to convert wind speed from CIRSOC 102 to ASCE 7 -05?

Thanks in advance.

MZ

RE: EBF Systems

I’m not certain but I think it has something to do with the location and formation of plastic hinges. It is preferable that the hinges form in the members not the connections, and further preferable that the hinges are in beams not columns. Beams can go into catenary and still support load, columns will collapse. If hinge occurs in member then may be possibility of repair, by replacing beam. If connection fails then may have to replace column.

My understanding of an EBF is that there is potential for formation of plastic hinge in column.

Regards

Conrad Harrison

B.Tech (mfg & mech), MIIE, gradTIEAust

mailto:sch.tectonic@bigpond.com

Adelaide

South Australia

From: rahman shahshenas [mailto:remish60@gmail.com]
Sent: Thursday, 12 June 2008 16:28
To: seaint@seaint.org
Subject: EBF Systems

Hello All,

In regard to Eccentrically Braced Frames, do you know any method for installing such braces at excisting R/C buildings, and why applying EBF at high gravity is unfavorable.

Regards
--
Rahman SHAHSHENAS

Bogazici University

Wednesday, June 11, 2008

EBF Systems

Re: Wind Load on Deep Snow Profile

One other thought, as the wind presure increases, will it blow the
snow off the roof? If the drift is parallel to the wind load, then
there is not an increase in area. This is the typical case considered
in the code. If the wind blows the other way, the drift is reduced.
The wind won't load the drift zone in the first place. Any drift
caused by wind blowing the other direction will be moved again by the
new wind direction.

This really needs a picture.

Jake Watson SE
Salt lake city, UT

On 6/10/08, Paul Blomberg <paul.blomberg@gmail.com> wrote:
> I'm putting together design criteria and am curious how you handle / take
> into consideration the added projected wind surface area due to built up
> snow on the roof of a building?
>
> I have a new metal building going into northwestern Montana and the ground
> snow load is over 120 psf. That is a relatively deep pile of snow. When
> you put together your load case that combines snow plus wind, do you
> increase the projected profile of the building to reflect the height of the
> snow on the roof and the associated wind load?
>
> I normally include this additional snow profile to calc the wind loads but
> this my first heavy snow load situation where the increased load is
> significant.
>
> Paul.
> Phoenix, AZ
>

******* ****** ****** ****** ******* ****** ****** ********

1985 pre-manufactured building

I am evaluating an existing pre-manufactured building that was designed
in 1985 under the 1982 UBC. The roof purlins are 12 gage. Is it safe
to assume that these are Fy = 50 ksi? At 33 ksi, they don't meet the
loads in the General Notes. Also, would the columns and frames be A36?
A572 Gr 50 was available then, but I am not sure if that material was
used by the manufacturers. Would it have been cheaper to use thicker
and wider plates of A36 versus some material with Fy = 50 ksi.

******* ****** ****** ****** ******* ****** ****** ********

Re: 1 or 2 piles under a cap

Jordan

David Maynard wrote:
> John Pack, PE, of IMR, Inc. out of Denver, CO, has done extensive studies
> with AB Chance Helical Piers. If you do an installation torque of 10,000
> ft. lbs, you could get an Ultimate Capacity of the pier to be 100 kips.
> With a factor of safety of 2.0, that's a 50 kips design capacity. If they
> have the materials to build a 50 kip (design capacity, 100 kip ultimate
> capacity) helical pier (heavier shaft, several helix flights, etc., then you
> can use a local helical pier for this particular load and isolate it from
> the rest of the foundation.
>
> Otherwise, with the spread that you have, you would have to design the
> foundation as a concrete beam between these two piers. I might not have a
> clear understanding of your problem. If you had a diagram to share, I might
> have more information.
>
> Dave Maynard
> Gillette, WY
>
>
> -----Original Message-----
> From: Jordan Truesdell, PE [mailto:seaint2@truesdellengineering.com]
> Sent: Wednesday, June 11, 2008 9:28 AM
> To: seaint@seaint.org
> Subject: 1 or 2 piles under a cap
>
> Does anyone have a rule of thumb for using less than 3 piles (helicals,
> in my case) under a pile cap? I've got a condition where I have some
> lightly loaded pile caps - about 45k - but their supporting pyramidal
> piers which have a 40" square base. The tops of the piers will support
> a beam line and concrete plank, so there's no lateral loads. The piers
> are 16-20' apart and are in a (roughly) straight line, so I can't
> efficiently combine multiple piers under a single footing.
>
>

******* ****** ****** ****** ******* ****** ****** ********

Re: Brittle and Ductile steel elements

On Jun 11, 2008, at 3:46 PM, Garner, Robert wrote:

> No, ductility is measured as at least 14% elongation and at least
> 30% reduction in area. I have a chart listing the following
> fasteners as ductile by this definition:
Not necessarily true either. Ductility is strongly affected by the
presence of a notch in stronger steels with %EL between 20% and 25%. %
elongation drops off some with temperature, but it doesn't drop off
as fast as Charpy test results.

Christopher Wright P.E. |"They couldn't hit an elephant at
chrisw@skypoint.com | this distance" (last words of Gen.
.......................................| John Sedgwick, Spotsylvania
1864)
http://www.skypoint.com/~chrisw/

******* ****** ****** ****** ******* ****** ****** ********

Re: Brittle and Ductile steel elements

On Jun 11, 2008, at 3:15 PM, Jnapd@aol.com wrote:

> In regard to anchors embedded in concrete ACI 318 App D regarding
> Brittle and Ductile steel elements
> Is it correct to assume if the Fu of the bolt is 65ksi or less then
> it is Ductile and over 65ksi is Brittle ??
No. You really need to know more about the material characteristics.
I don't know what the Code says maybe they have their own definition
of ductile and brittle, but UTS alone doesn't determine brittle
behavior. In carbon steels brittle behavior is strongly temperature
dependent, but that depends on processing variables. Carbon and
manganese content also have a strong effect. Carbon content over
0.45% makes for more brittle behavior as does manganese content below
about 0.60%. Quenched and tempered steel is more brittle than
annealed material of the same chemistry. It's not a topic with a few
quick rules of thumb.

******* ****** ****** ****** ******* ****** ****** ********

RE: 1 or 2 piles under a cap

John Pack, PE, of IMR, Inc. out of Denver, CO, has done extensive studies
with AB Chance Helical Piers. If you do an installation torque of 10,000
ft. lbs, you could get an Ultimate Capacity of the pier to be 100 kips.
With a factor of safety of 2.0, that's a 50 kips design capacity. If they
have the materials to build a 50 kip (design capacity, 100 kip ultimate
capacity) helical pier (heavier shaft, several helix flights, etc., then you
can use a local helical pier for this particular load and isolate it from
the rest of the foundation.

Otherwise, with the spread that you have, you would have to design the
foundation as a concrete beam between these two piers. I might not have a
clear understanding of your problem. If you had a diagram to share, I might
have more information.

Dave Maynard
Gillette, WY

-----Original Message-----
From: Jordan Truesdell, PE [mailto:seaint2@truesdellengineering.com]
Sent: Wednesday, June 11, 2008 9:28 AM
To: seaint@seaint.org
Subject: 1 or 2 piles under a cap

Does anyone have a rule of thumb for using less than 3 piles (helicals,
in my case) under a pile cap? I've got a condition where I have some
lightly loaded pile caps - about 45k - but their supporting pyramidal
piers which have a 40" square base. The tops of the piers will support
a beam line and concrete plank, so there's no lateral loads. The piers
are 16-20' apart and are in a (roughly) straight line, so I can't
efficiently combine multiple piers under a single footing.

--
Jordan

******* ****** ****** ****** ******* ****** ****** ********

RE: Brittle and Ductile steel elements

Return Receipt

Your RE: Brittle and Ductile steel elements
document:

was Tom Hunt/AV/FD/FluorCorp
received
by:

at: 06/11/2008 14:41:00 PDT

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If you are not the intended recipient of this message you are
hereby notified that any use, review, retransmission, dissemination,
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Any views expressed in this message are those of the individual
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RE: Brittle and Ductile steel elements

"PCA Notes on ACI 318-05 Building Code Requirements for Structural
Concrete with Design Applications " see
http://www.cement.org/bookstore/profile.asp?store=&pagenum=1&pos=0&catID
=&id=8665

Wesley C. Werner

-----Original Message-----
From: Drew Morris [mailto:dmorris@bbfm.com]
Sent: Wednesday, June 11, 2008 5:06 PM
To: seaint@seaint.org
Subject: Re: Brittle and Ductile steel elements

Wesley Werner wrote:
> No. All ASTM F1554 anchor rods are ductile and the highest Fu is
> 125 ksi for F1554, Gr 105. Look under the definitions in the front of
> Appendix D for the requirements for ductile vs brittle. The
> requirements relate to elongation and reduction in area under tension
> loading. PCA's commentary on 318-05 Appendix D has a chart with these
> properties listed for some common steel types. If you are just
> beginning to work with Appendix D, I would highly recommend PCA's
> commentary on the code. It has many helpful resources.
>
>
> /Wesley C. Werner, PE/
>
Which PCA publication is this?

******* ****** ****** ****** ******* ****** ****** ********

Re: Brittle and Ductile steel elements

******* ****** ****** ****** ******* ****** ****** ********

RE: Brittle and Ductile steel elements

No. All ASTM F1554 anchor rods are ductile and the highest Fu is 125 ksi for F1554, Gr 105. Look under the definitions in the front of Appendix D for the requirements for ductile vs brittle. The requirements relate to elongation and reduction in area under tension loading. PCA's commentary on 318-05 Appendix D has a chart with these properties listed for some common steel types. If you are just beginning to work with Appendix D, I would highly recommend PCA's commentary on the code. It has many helpful resources.

Wesley C. Werner, PE

From: Jnapd@aol.com [mailto:Jnapd@aol.com]
Sent: Wednesday, June 11, 2008 4:16 PM
To: seaint@seaint.org
Subject: Brittle and Ductile steel elements

Hello All

In regard to anchors embedded in concrete ACI 318 App D regarding Brittle and Ductile steel elements

Is it correct to assume if the Fu of the bolt is 65ksi or less then it is Ductile and over 65ksi is Brittle ??

Joe Venuti
Johnson & Nielsen Associates
Palm Springs, CA

Vote for your city's best dining and nightlife. City's Best 2008.

RE: Brittle and Ductile steel elements

No, ductility is measured as at least 14% elongation and at least 30% reduction in area. I have a chart listing the following fasteners as ductile by this definition:

ASTM A307 A or B

ASTM A 354 BC or BD

ASTM A 449 Type 1

ASTM A 687

ASTM F 1554 GR 36, 55 or 105

Bob Garner, S.E.

From: Jnapd@aol.com [mailto:Jnapd@aol.com]
Sent: Wednesday, June 11, 2008 1:16 PM
To: seaint@seaint.org
Subject: Brittle and Ductile steel elements

Hello All

In regard to anchors embedded in concrete ACI 318 App D regarding Brittle and Ductile steel elements

Is it correct to assume if the Fu of the bolt is 65ksi or less then it is Ductile and over 65ksi is Brittle ??

Joe Venuti
Johnson & Nielsen Associates
Palm Springs, CA

Vote for your city's best dining and nightlife. City's Best 2008.

Brittle and Ductile steel elements

Hello All

In regard to anchors embedded in concrete ACI 318 App D regarding Brittle and Ductile steel elements

Is it correct to assume if the Fu of the bolt is 65ksi or less then it is Ductile and over 65ksi is Brittle ??

Joe Venuti
Johnson & Nielsen Associates
Palm Springs, CA

Vote for your city's best dining and nightlife. City's Best 2008.

Out-of-plane anchorage

List,

There used to be a requirement that wood elements resisting out-of-plane forces from CMU or concrete walls had to be a minimum of 2-1/2” thick. This does not seem to be a part of ASCE 7-05. Can someone please confirm that this requirement does not exist in the ASCE 7-05? This requirement is in the 2001 California Building Code, so it might never have been in the ASCE, but I just want to check.

TIA,

Doug Mayer, SE

Structural Engineer

TaylorTeter

Partnership

Visit our new website at www.taylorteter.com

Re: A little help, please!

It says "permissible" so I understand that you can use the 2 times the
drift criteria as long as this diaphragm deflection amount does not
exceed the permissible deflection of the attached elements, so
therefore I don't think it is contradictory.

On Wed, Jun 11, 2008 at 2:18 PM, <ECVAl3@aol.com> wrote:
> I'm a bit confused. Could someone clarify this for me, please.
>
> The IBC Section 2305.2.1 states: "Wood diaphragms are permitted to be used
> to resist horizontal forces
> provided the deflection in the plane of the diaphragm, as determined by
> calculations..., does not exceed the permissible deflection of attached
> distributing or resisting elements..."
>
> But the ASCE Section 12.3.1.3 Calculated Flexible Diaphragm Condition,
> states: "Diaphragms...are permitted to be idealized as flexible where the
> computed maximum in-plane deflection..is more than 2 times the average story
> drift of adjoining vertical elements of the seismic force-resisting
> system..."
>
> Are these contradictory requirements or am I reading them wrong?
>
> S.Macie. P.E.
> SLO, CA
>
>
> ________________________________
> Vote for your city's best dining and nightlife. City's Best 2008.

******* ****** ****** ****** ******* ****** ****** ********

A little help, please!

I'm a bit confused. Could someone clarify this for me, please.

The IBC Section 2305.2.1 states: "Wood diaphragms are permitted to be used to resist horizontal forces

provided the deflection in the plane of the diaphragm, as determined by calculations..., does not exceed the permissible deflection of attached distributing or resisting elements..."

But the ASCE Section 12.3.1.3 Calculated Flexible Diaphragm Condition, states: "Diaphragms...are permitted to be idealized as flexible where the computed maximum in-plane deflection..is more than 2 times the average story drift of adjoining vertical elements of the seismic force-resisting system..."

Are these contradictory requirements or am I reading them wrong?

S.Macie. P.E.

SLO, CA

Vote for your city's best dining and nightlife. City's Best 2008.

1 or 2 piles under a cap

--
Jordan

******* ****** ****** ****** ******* ****** ****** ********

RE: Wind Load on Deep Snow Profile

Paul,

That’s the first I’ve ever heard of the concept of increasing wind loads due to built-up snow. Although I will grant most of my experience is mid-Atlantic US, 20-30psf ground snow loads.

I can’t see it making much of a difference. It doesn’t seem to me, even with heavy snow buildup, that you’d increase the mean roof height much more than 5 feet or so. That translates (glancing at the height adjustment tables) to about a 5% increase in the wind pressure. But the wind load is nowhere near that accurate to begin with. For starters, if you’re using the low-rise method, you’re locked to the design pressures for 30 feet mean roof height even if you only have a 15 or 20 foot mean roof height. Then if you’re in Exposure B, you’re actually designing to almost an Exposure C surface roughness; the roughness values picked for Exposure B are at the lower bound (i.e. nearly open terrain).

Plus, if you’ve got that much snow it’s probably building up on the ground as well as on the roof too. So “grade plane” is higher too…

Regards,

Gary

Gary J. Ehrlich, PE
Program Manager, Structural Codes & Standards
National Association of Home Builders (NAHB)
1201 15th Street, NW, Washington, DC 20005
ph: 202-266-8545 or 800-368-5242 x8545
fax: 202-266-8369
gehrlich@nahb.com

From: Paul Blomberg [mailto:paul.blomberg@gmail.com]
Sent: Tuesday, June 10, 2008 2:44 PM
To: seaint
Subject: Wind Load on Deep Snow Profile

I'm putting together design criteria and am curious how you handle / take into consideration the added projected wind surface area due to built up snow on the roof of a building?

I have a new metal building going into northwestern Montana and the ground snow load is over 120 psf. That is a relatively deep pile of snow. When you put together your load case that combines snow plus wind, do you increase the projected profile of the building to reflect the height of the snow on the roof and the associated wind load?

I normally include this additional snow profile to calc the wind loads but this my first heavy snow load situation where the increased load is significant.

Paul.

Phoenix, AZ

Re: 30" wide catwalk

Julius,
Just to add a thought to all the others on this subject. Up here, our
electrical code requires that where ever there is an electrical panel,
that there must be at least 30 inches clear in front of the closest
electrical component. I am just going by memory but that is the gist of it.
Gary

David Topete wrote:
> Julius,
> I would agree that the catwalk that you are working on can be 24" wide
> if it is considered access for maintenance of the conveying
> equipment. Essentially, you should be able to argue that the expected
> occupant load is no more than maybe two "workers" along the length of
> catwalk. And, it's an industrial facility so I strongly doubt that
> this catwalk is to be ADA compliant. Good luck.
>
> List:
>
> Can anyone confirm that the *minimum width of catwalk* used in
> heavy industrial (i.e. for Bulk/Material handling conveyor system)
> has been changed from *24" to 30" wide?* One of our independent
> checkers made comments regarding this change.
>
> I checked with IBC 2000, 2003 and 2006 nothing mentioned about the
> change.
>
> Thanks,
>
> Julius
>
> *Engr. Julius Micayas*
> *P.E. license no.32969*
> *Project Manager/Sr Lead Structural Engineer*
> */River Consulting LLC/*
> *111 Veterans Memorial Boulevard*
> * /Suite 1600/*
> */Metairie, Louisiana 70005/*
> *Phone - 504-841-3014 (direct)*
> *504 837-5275 (office)*
> *Fax - 504-837-2986*
> *E-mail: *_jmicayas@riverconsulting.com_
> <mailto:jmicayas@riverconsulting.com>
> *W-page: *_www.riverconsulting.com_
> <http://www.riverconsulting.com/>
>
>
> ------------------------------------------------------------
> The information transmitted is intended only for the person
> or entity to which it is addressed and may contain
> proprietary, business-confidential and/or privileged material.
> If you are not the intended recipient of this message you are
> hereby notified that any use, review, retransmission, dissemination,
> distribution, reproduction or any action taken in reliance upon
> this message is prohibited. If you received this in error, please
> contact the sender and delete the material from any computer.
>
> Any views expressed in this message are those of the individual
> sender and may not necessarily reflect the views of the company.
> ------------------------------------------------------------
>
>
>
>
>
> --
> David Topete, SE

******* ****** ****** ****** ******* ****** ****** ********

Tuesday, June 10, 2008

Thank You

List,

I personally thank you for all the responses received.

Best regards,

Julius

Engr. Julius Micayas

P.E. license no.32969

Project Manager/Sr Leasd Structural Engineer

River Consulting LLC

111 Veterans Memorial Boulevard

Suite 1600

Metairie, Louisiana 70005

Phone - 504-841-3014 (direct)

504 837-5275 (office)

Fax - 504-837-2986

E-mail: jmicayas@riverconsulting.com

W-page: www.riverconsulting.com

RE: China EQ

Sorry, I meant 20 seconds.
[Some reason my correction post posted some hours ago did not appear on the
list]

-Suresh

-----Original Message-----
From: Ralph Kratz [mailto:rhkratzse@aol.com]
Sent: Tuesday, June 10, 2008 1:59 PM
To: <seaint@seaint.org>
Cc: seaint@seaint.org
Subject: Re: China EQ

Hmmm, Suresh, I was on the roof of my house in Richmond & Loma Prieta seemed
to last all of about 30 seconds, altho I recognize that imperceptable
shaking may have lasted much longer. I thought the length is typically
given as about 5 minutes.

Ralph

Sent from my iPhone

On Jun 10, 2008, at 11:42 AM, Drew Morris <dmorris@bbfm.com> wrote:

> Acharya, Suresh wrote:
>> Worth adding: Sichuan earthquake lasted several minutes compared to
>> Loma Prieta earthquake which lasted for 20 minutes. Both occurred
>> at similar depths (18 km).
>> Suresh Acharya, S.E.
> The USGS gave the duration of the Loma Preita EQ as being on the
> order of 20 seconds. Does the 20 minutes you quote include
> aftershocks?
>
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