Structural Engineering: 2008-07-27

Saturday, August 2, 2008

Re: EQ and international codes

Yup, sort of like "compassionate cons..."

Oops.

In a message dated 8/2/08 5:48:12 AM, ghodgson@bellnet.ca writes:

Isn't corrupt dictator an oxymoron?

Jnapd@aol.com wrote:
> Is that like a totally corrupt dictator or dictator by council.
>
>
> There was an article in this morning's paper about China. It said it is
> neither a communist country nor a capitalist country but combines the
> worst features of both.
> Gary
>
> Andrew Kester, P.E. wrote:
> > And in China who knows what happens in a communist country, if someone
> > wants something built fast and cheap, they can probably steamroll it
> > through. There are not lawyers on every corner either, and a court
> > system that will support legal claims....
> >
> >
>
> Joe Venuti
> Johnson & Nielsen Associates
> Palm Springs, CA
>

**************
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Re: EQ and international codes

Isn't corrupt dictator an oxymoron?

Jnapd@aol.com wrote:
> Is that like a totally corrupt dictator or dictator by council.
>
>
> There was an article in this morning's paper about China. It said it is
> neither a communist country nor a capitalist country but combines the
> worst features of both.
> Gary
>
> Andrew Kester, P.E. wrote:
> > And in China who knows what happens in a communist country, if someone
> > wants something built fast and cheap, they can probably steamroll it
> > through. There are not lawyers on every corner either, and a court
> > system that will support legal claims....
> >
> >
>
> Joe Venuti
> Johnson & Nielsen Associates
> Palm Springs, CA
>
>
>
> ------------------------------------------------------------------------
> Get fantasy football with free live scoring. Sign up for FanHouse
> Fantasy Football today
> <http://www.fanhouse.com/fantasyaffair?ncid=aolspr00050000000020>.

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Friday, August 1, 2008

Re: ACI 318 App D, and wedge anchors - More confusion [NOT A JOKE!!]

In a message dated 8/1/08 12:44:52 PM, T.W.Allen@cox.net writes:

The more I read, the more confused I become.

Guess that just proves that you're getting senile, probably incapable of practicing engineering any longer. [JOKE!!!]

OR: Does it prove that the code-writers, whoever in hail they are, are completely incompetent to write a code that can be readily understood and interpreted by a run-of-the-mill engineer who has *only* 5 or 6 years of university plus a bunch of years of experience and many days of testing, but unfortunately may not have a Ph.D. in engineering research and an unlimited client-supported budget for research disguised as design. [NOT A JOKE!!!!]

Really.

Sincerely,

Ralph

**************
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RE: ACI 318 App D, and wedge anchors - More confusion

Suppose I determine that a particular post installed anchor has an ultimate shear capacity of 1,640 lbs based on ACI Appendix D. The anchor is in a high seismic area and the anchor is not ductile.

1. How are the 0.75 (ACI D.3.3.3) and 2.5 (CBC 1908.1.16) factors combined for seismic loads?

Would the allowable capacity for seismic loading be (0.75)(1,640/1.4)/2.5 = 351 lbs?

The 0.75 factor is due to ACI D.3.3.3 and the 2.5 factor is due to 2007 CBC section 1908.1.16.

The way I read 1908.1.16, it is “either/or” relative to the 0.75 and the 2.5 factors.

In other words, if the anchor is ductile, would the (ASD) capacity be (0.75)(1,640/1.4) = 879 lbs.?

If the anchor is not ductile, would the (ASD) capacity be (1,640/1.4)/2.5 = 469 lbs?

2. Are these factors still used when comparing the capacity of the anchor to the demand induced by wind loading?

In other words, would the allowable capacity of the anchor in resisting wind loads be (1,640/1.6) = 1,025 lbs?

(note the absence of the 0.75 and 2.5 factors).

The more I read, the more confused I become.

T. William (Bill) Allen, S.E.

ALLEN DESIGNS

Consulting Structural Engineers
V (949) 248-8588 • F(949) 209-2509

Re: Overstrength factor

Rahman,

I wouldn't assume anything with software. They have been known to have some mistakes from time to time :-)

Verify with some hand calcs (not too tough on steel code checking) till you get the confidence you need to use the tool.

-gm

On Thu, Jul 31, 2008 at 11:30 PM, rahman shahshenas <remish60@gmail.com> wrote:

Thank you Gerard

I asked CSI too, and the response was to chk "design manual". I think we should assume that the program chk for it even though we can not see any report.

RS

On Thu, Jul 31, 2008 at 6:31 PM, Gerard Madden, SE <gmse4603@gmail.com> wrote:

Yes Etabs checks it,

unfortunately in many cases, it doesn't let you know it checks this unless it doesn't work, then you'll see the red message.

I've asked CSI to consider placing this check in the output display in future releases so the program is more transparent.

I had it happen in a design recently that had a lot of discontinuous frames that transfered out load through the first floor diaphragm but continued down to a basement. The program didn't explicitly say it was checking for omega forces on the columns... but when you make them too small, it will tell you there is a problem.

-gm

On Thu, Jul 31, 2008 at 1:59 AM, rahman shahshenas <remish60@gmail.com> wrote:

Dear All,

In regard to program defults at SAP2000 or Etabs, I design a steel structure (concentrically braced) and need to know the program checks for (1.2 + 0.2SSDS) DL + 1.0 LL ± Ehm. I know that this chk is required in special seismic cases. So I followed the design manual and found out that, this combination is created as internal combination automatically. however I couldn't see any report upon that claim. so does it really chk for those combinations?

thanks in advance

--
Rahman SHAHSHENAS

Orient Research Consulting Engineers
Akatlar-Istanbul

--
Rahman SHAHSHENAS

Bogazici University

Re: EQ and international codes

Is that like a totally corrupt dictator or dictator by council.

There was an article in this morning's paper about China. It said it is
neither a communist country nor a capitalist country but combines the
worst features of both.
Gary

Andrew Kester, P.E. wrote:
> And in China who knows what happens in a communist country, if someone
> wants something built fast and cheap, they can probably steamroll it
> through. There are not lawyers on every corner either, and a court
> system that will support legal claims....
>
>

Joe Venuti
Johnson & Nielsen Associates
Palm Springs, CA

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FW: Slab on Grade Detail question

First message bounced, here it is.

From: Christopher Banbury
Sent: Friday, August 01, 2008 10:11 AM
To: 'seaint@seaint.org'
Subject: RE: Slab on Grade Detail question

I think that is a jurisdiction by jurisdiction thing. The same issue exists when using a stemwall or foundation wall footing in that the footing extends out past the face of the wall. Most jurisdictions here measure setbacks from the face of the wall, not from the footing or even the overhang, up to a certain maximum overhang like 24”.

Christopher Banbury, PE

President

Ark Engineering, Inc.

PO Box 10129, Brooksville, FL 34603

22 North Broad ST, Brooksville, FL 34601

Phone: (352) 754-2424

Fax: (352) 754-2412

www.arkengineering.net

RE: footing edges

I’m sorry, I misunderstood, if the foundation wall will extend into the setback above grade then I think the AHJs here might measure the setback from the footing.

Christopher Banbury, PE

President

Ark Engineering, Inc.

PO Box 10129, Brooksville, FL 34603

22 North Broad ST, Brooksville, FL 34601

Phone: (352) 754-2424

Fax: (352) 754-2412

www.arkengineering.net

From: Dennis Wish [mailto:dennis.wish@verizon.net]
Sent: Thursday, July 31, 2008 8:00 PM
To: seaint@seaint.org
Subject: RE: footing edges

I’m sorry – I did not state it correctly. There is a 5’-0” setback on the side-yards. This is, I assume, from the face of the exterior wall of the home to the property line. In the past, the building department would only allow the roof overhang to encroach into the setback by up to 18-inches. My question had to do with the building foundation. While the exterior wall is exactly 5’-0” from the property line, I want to offset the turn-down edge of a slab on grade (similar to the base of a “T” shaped foundation) and have the edge of the concrete footing 3-1/2” into the 5’-0” setback. I don’t believe this is a problem, but wanted to verify if this is the case or if there is any specific section of the IBC 2006 that would restrict this encroachment?

Thanks Bob,

Dennis

Re: EQ and international codes

There was an article in this morning's paper about China. It said it is
neither a communist country nor a capitalist country but combines the
worst features of both.
Gary

Andrew Kester, P.E. wrote:
> And in China who knows what happens in a communist country, if someone
> wants something built fast and cheap, they can probably steamroll it
> through. There are not lawyers on every corner either, and a court
> system that will support legal claims....
>
>

Thursday, July 31, 2008

Re: Overstrength factor

Thank you Gerard

I asked CSI too, and the response was to chk "design manual". I think we should assume that the program chk for it even though we can not see any report.

RS

On Thu, Jul 31, 2008 at 6:31 PM, Gerard Madden, SE <gmse4603@gmail.com> wrote:

Yes Etabs checks it,

unfortunately in many cases, it doesn't let you know it checks this unless it doesn't work, then you'll see the red message.

I've asked CSI to consider placing this check in the output display in future releases so the program is more transparent.

I had it happen in a design recently that had a lot of discontinuous frames that transfered out load through the first floor diaphragm but continued down to a basement. The program didn't explicitly say it was checking for omega forces on the columns... but when you make them too small, it will tell you there is a problem.

-gm

On Thu, Jul 31, 2008 at 1:59 AM, rahman shahshenas <remish60@gmail.com> wrote:

Dear All,

In regard to program defults at SAP2000 or Etabs, I design a steel structure (concentrically braced) and need to know the program checks for (1.2 + 0.2SSDS) DL + 1.0 LL ± Ehm. I know that this chk is required in special seismic cases. So I followed the design manual and found out that, this combination is created as internal combination automatically. however I couldn't see any report upon that claim. so does it really chk for those combinations?

thanks in advance

--
Rahman SHAHSHENAS

Orient Research Consulting Engineers
Akatlar-Istanbul

--
Rahman SHAHSHENAS

Bogazici University

Re: Overstrength factor

Thank you Gerard

I asked CSI too, and the response was to chk "design manual". I think we should assume that the program chk for it even though we can not see any report.

RS

On Thu, Jul 31, 2008 at 6:31 PM, Gerard Madden, SE <gmse4603@gmail.com> wrote:

Yes Etabs checks it,

unfortunately in many cases, it doesn't let you know it checks this unless it doesn't work, then you'll see the red message.

I've asked CSI to consider placing this check in the output display in future releases so the program is more transparent.

I had it happen in a design recently that had a lot of discontinuous frames that transfered out load through the first floor diaphragm but continued down to a basement. The program didn't explicitly say it was checking for omega forces on the columns... but when you make them too small, it will tell you there is a problem.

-gm

On Thu, Jul 31, 2008 at 1:59 AM, rahman shahshenas <remish60@gmail.com> wrote:

Dear All,

In regard to program defults at SAP2000 or Etabs, I design a steel structure (concentrically braced) and need to know the program checks for (1.2 + 0.2SSDS) DL + 1.0 LL ± Ehm. I know that this chk is required in special seismic cases. So I followed the design manual and found out that, this combination is created as internal combination automatically. however I couldn't see any report upon that claim. so does it really chk for those combinations?

thanks in advance

--
Rahman SHAHSHENAS

Orient Research Consulting Engineers
Akatlar-Istanbul

--
Rahman SHAHSHENAS

Bogazici University

Re: Slab on Grade Detail question

My experience with setbacks is that they apply only to walls above ground level. Thus footings may extend into this setback.

Stan Scholl, P.E.

Laguna Beach, CA

Re: footing edges

Those are usually privacy and daylighting setbacks. You should be okay with sub-grade work unless there is a utility easement.

-gm

On Thu, Jul 31, 2008 at 5:03 PM, <ECVAl3@aol.com> wrote:

I believe the original intent of the setback was to allow access for emergencies and help prevent fire from easily spreading to adjacent structures. I don't see how a footing below ground could interfere with access. I have designed many a footing that encroached into the setback without resulting in a plan correction.

S.Macie, SLO

I'm sorry – I did not state it correctly. There is a 5'-0" setback on the side-yards. This is, I assume, from the face of the exterior wall of the home to the property line. In the past, the building department would only allow the roof overhang to encroach into the setback by up to 18-inches. My question had to do with the building foundation. While the exterior wall is exactly 5'-0" from the property line, I want to offset the turn-down edge of a slab on grade (similar to the base of a "T" shaped foundation) and have the edge of the concrete footing 3-1/2" into the 5'-0" setback. I don't believe this is a problem, but wanted to verify if this is the case or if there is any specific section of the IBC 2006 that would restrict this encroachment?

Thanks Bob,

Dennis

From: Bob Freeman [mailto:robert.freeman@idsse.com]
Sent: Thursday, July 31, 2008 3:52 PM
To: seaint@seaint.org
Subject: footing edges

Hi Dennis:

If I understand your question correctly, the edge of the t-shaped footing should end precisely at the property line, unless your client wants to purchase adjacent lands.

When you mention setback lines, I think of zoning setbacks. (Sometimes 6 feet from the property line, depending on your location.) It is acceptable to extend past a zoning setback line with a footing under ground, as long as face of the building wall is no closer to the property than zoning allows. Sometimes municipal codes allow overhangs to extend into zoning setbacks.

Respectfully,

Bob Freeman, AIA, EIT

IDS Group, Inc.

(949) 387-8500

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Re: Prestressed Double T Camber

Wontae,

There was a theory many years ago that the Youngs Modulus does not increase as quickly as compressive strength, especially in cases like steam curing. I remember they were talking about a Maturity Coefficient. Have not heard of it since but it would explain your dilemma. With the rapid strength increase under steam curing, the modulus is probably lagging. Because you are leaving these planks for 3 days instead of 8 hours there may be a large increase in Ec over that time. The computer program would not be predicting that.

At 10:07 PM 25/07/2008, you wrote:

Thank you for your responses!

I did not know that late stripping of prestressed members from the mold would cause severe camber difference.
Though, this deviation is within product tolerance (0.25"/10ft , max 0.75"/10ft) according to PCI Design Handbook;
I got 2" while 0.75x60'/10'=.45", which is very big number!

I understand camber is related with E and E is eventually affected by f'c.
In my case, when I ran a program to compare how f'c affects camber using 3500psi (@stripping)/5000psi(@28-day)
and 6000psi(@stripping)/8000psi(@28-day), cambers were only 0.5" difference.
I am still puzzled at what caused 2" camber difference.
In most precast plants, steam curing is popular and I think concrete strength does not change much between 12 hours
and 36 hours after concrete placement
because like my company weekend concrete mix may be used and steam is not provided during weekend.

Here is my conclusion:
we have to cut strands if prestressed members stay in the mold for extended hours in order to prevent strand failure and/or concrete spalling
and excessive camber variance.
I did not know second part was so important.

Thanks!

PS: Jim G., I appreciate your remedial advice!

--
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Regards Gil Brock
Prestressed Concrete Design Consultants Pty. Ltd. (ABN 84 003 163 586)
5 Cameron Street Beenleigh Qld 4207 Australia
Ph +61 7 3807 8022               Fax +61 7 3807 8422
email:            gil@raptsoftware.com
email:            sales@raptsoftware.com
email:            support@raptsoftware.com
webpage:         http://www.raptsoftware.com

Re: footing edges

I believe the original intent of the setback was to allow access for emergencies and help prevent fire from easily spreading to adjacent structures. I don't see how a footing below ground could interfere with access. I have designed many a footing that encroached into the setback without resulting in a plan correction.

S.Macie, SLO

I'm sorry – I did not state it correctly. There is a 5'-0" setback on the side-yards. This is, I assume, from the face of the exterior wall of the home to the property line. In the past, the building department would only allow the roof overhang to encroach into the setback by up to 18-inches. My question had to do with the building foundation. While the exterior wall is exactly 5'-0" from the property line, I want to offset the turn-down edge of a slab on grade (similar to the base of a "T" shaped foundation) and have the edge of the concrete footing 3-1/2" into the 5'-0" setback. I don't believe this is a problem, but wanted to verify if this is the case or if there is any specific section of the IBC 2006 that would restrict this encroachment?

Thanks Bob,

Dennis

From: Bob Freeman [mailto:robert.freeman@idsse.com]
Sent: Thursday, July 31, 2008 3:52 PM
To: seaint@seaint.org
Subject: footing edges

Hi Dennis:

If I understand your question correctly, the edge of the t-shaped footing should end precisely at the property line, unless your client wants to purchase adjacent lands.

When you mention setback lines, I think of zoning setbacks. (Sometimes 6 feet from the property line, depending on your location.) It is acceptable to extend past a zoning setback line with a footing under ground, as long as face of the building wall is no closer to the property than zoning allows. Sometimes municipal codes allow overhangs to extend into zoning setbacks.

Respectfully,

Bob Freeman, AIA, EIT

IDS Group, Inc.

(949) 387-8500

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RE: footing edges

Thanks Bob,

Dennis

From: Bob Freeman [mailto:robert.freeman@idsse.com]
Sent: Thursday, July 31, 2008 3:52 PM
To: seaint@seaint.org
Subject: footing edges

Hi Dennis:

If I understand your question correctly, the edge of the t-shaped footing should end precisely at the property line, unless your client wants to purchase adjacent lands.

When you mention setback lines, I think of zoning setbacks. (Sometimes 6 feet from the property line, depending on your location.) It is acceptable to extend past a zoning setback line with a footing under ground, as long as face of the building wall is no closer to the property than zoning allows. Sometimes municipal codes allow overhangs to extend into zoning setbacks.

Respectfully,

Bob Freeman, AIA, EIT

IDS Group, Inc.

(949) 387-8500

Re: ACI 318 App D, and wedge anchors

Jared,

What brittle failure can we/they talk about with the safety factors per Appendix D?

Some normal structure (say, a house) is designed with an overall safety factor of, say, about 3 - against its actual physical failure. This means that it will take an unheard of wind or earthquake to take the structure down (cause structural failure). This further means that long before any (brittle or not) kind of failure in the anchors, the structure itself will be ruined, and the performance of the anchor will not matter.

The logic (or lack thereof) of the situation leads to the conclusion that the structure itself represents a "weak link." The failure of this "link" will inevitably come earlier than that of the anchor, practically assuring that the anchor will never fail.

Besides, there are no reported epoxy anchor failures. None.

V. Steve Gordin, SE
Irvine CA

----- Original Message -----

From: Jared Keyser

To: Seaint@seaint.org

Sent: Thursday, July 31, 2008 15:55

Subject: RE: ACI 318 App D, and wedge anchors

Bill,

You are right that the ductility issue applies in both tension and shear. But if you add enough anchors far enough down on the wall, you can get the concrete connection strength of the multiple anchors to be greater than the single threaded rod in tension to the hold down at top of the bracket. This was an admittedly over-kill design to meet the requirement for a single location, so it was feasible. I wouldn't recommend it to be used at multiple locations in a retrofit, since it would get real expensive quick.

As far as the series of calcs goes, I ran ACI appendix D calcs in accordance with the ICC ES report for a number of anchor sizes and different "seismically approved" anchors and quickly realized that is was not possible to get the anchors to fail in a ductile manner in tension. I emailed Hilti tech support since these were the "seismically approved" anchors to see what I was missing. The tech support provided the following response in May 2007, in the middle of the post-installed anchor crisis:

"You are absolutely right and you are not missing anything. In most of the design cases concrete failure modes are governing and resulting in a brittle failure."

Hilti went on to provide information on the changes approved for ACI 2008 and the 2.5 factor for non-ductile designs. Combined with the lower values developed under the new testing criteria, this is unfortunately quite a burden on the post-installed anchors and in many cases makes them very difficult to be utilized and meet the direct wording of the code, especially for single anchor bolts or hold downs. I suspect that in a couple years, Hilti, Simpson and others will find a way to address this issue through new product developments that will meet the intent of the current codes.   In the mean time, I cringe every time I get the call from the contractor with a mis-aligned anchor bolt and the request to pop in a post-installed anchor.

Jared

The ductility issue (1/2.5) not only applies to tension anchors but shear
anchors as well.

Just curious, what "series of calcs" did you go through? Section D.3.3.5
isn't clear on this.

T. William (Bill) Allen, S.E.
ALLEN DESIGNS
Consulting Structural Engineers

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RE: ACI 318 App D, and wedge anchors

Bill,

You are right that the ductility issue applies in both tension and shear. But if you add enough anchors far enough down on the wall, you can get the concrete connection strength of the multiple anchors to be greater than the single threaded rod in tension to the hold down at top of the bracket. This was an admittedly over-kill design to meet the requirement for a single location, so it was feasible. I wouldn't recommend it to be used at multiple locations in a retrofit, since it would get real expensive quick.

As far as the series of calcs goes, I ran ACI appendix D calcs in accordance with the ICC ES report for a number of anchor sizes and different "seismically approved" anchors and quickly realized that is was not possible to get the anchors to fail in a ductile manner in tension. I emailed Hilti tech support since these were the "seismically approved" anchors to see what I was missing. The tech support provided the following response in May 2007, in the middle of the post-installed anchor crisis:

"You are absolutely right and you are not missing anything. In most of the design cases concrete failure modes are governing and resulting in a brittle failure."

Hilti went on to provide information on the changes approved for ACI 2008 and the 2.5 factor for non-ductile designs. Combined with the lower values developed under the new testing criteria, this is unfortunately quite a burden on the post-installed anchors and in many cases makes them very difficult to be utilized and meet the direct wording of the code, especially for single anchor bolts or hold downs. I suspect that in a couple years, Hilti, Simpson and others will find a way to address this issue through new product developments that will meet the intent of the current codes. In the mean time, I cringe every time I get the call from the contractor with a mis-aligned anchor bolt and the request to pop in a post-installed anchor.

Jared

The ductility issue (1/2.5) not only applies to tension anchors but shear
anchors as well.

Just curious, what "series of calcs" did you go through? Section D.3.3.5
isn't clear on this.

T. William (Bill) Allen, S.E.
ALLEN DESIGNS
Consulting Structural Engineers

footing edges

Hi Dennis:

If I understand your question correctly, the edge of the t-shaped footing should end precisely at the property line, unless your client wants to purchase adjacent lands.

Respectfully,

Bob Freeman, AIA, EIT

IDS Group, Inc.

(949) 387-8500

Slab on Grade Detail question

I am modifying my standard cross section of a SOG detail to have a portion of the footing at the base of the turn down edge protrude out from what I normally show as a straight vertical side of the slab from bottom of foundation to the top of slab. I want to use a 2x4 to extend the lower 8-inch of the turn down edge to form a lip for two reasons:

1. When setting flatwork (walkways, patio slabs etc.) against an existing SOG, the lip helps to prevent differential settlement over time by adding support for the flatwork.

2. When remodeling a foundation and attempting to use a proprietary shear element, the edge distance between the center of the anchor and embedment of the anchor into an epoxies hole will reduce the capacity for uplift on a traditional straight vertical turn down edge. This would allow for greater capacity and compliance with most adhesive manufacturers minimum clearance between the center of the threaded rod and the edge of the concrete foundation.

My question has to do with property setbacks. I’ve designed “L” shaped property walls and the foundation of the CMU wall is allowed on the clients side of the property line only. If the 3-1/2” protrusion of the bottom of the SOG of the home extends into the setback does it violate the property setback clearance requirements established by local municipalities or is the encroachment allowed as it is with the roof overhang?

Thanks in advance

Dennis S. Wish, PE

California Professional Engineer

Structural Engineering Consultant

La Quinta, CA 92253

dennis.wish@verizon.net

http://structuralist.wordpress.com

http://www.structuralist.net

RE: MWFRS Method 1 Question

No, that was the reason I was asking. Ben answered it as I had expected; no limit to stories but height limited to 60-feet. This makes a big difference. Breyer makes it clear that the windows and doors can be ignored as long as they work for cladding and can be closed during a high wind event. This was one of those questions that I originally had about garages in which homeowners leave their garage doors open. I also had questions related to homes with patio covers that are exposed on at least three sides and Breyer explained this as taking the entire structure into consideration while the point being based on securing the living area against the wind force driven debris.

Breyer’s book helped me a great detail and while I would include Method 2 in my Multi-Lat™ 2008 software it appears that for most Lateral Force Resisting Systems where seismic may govern, there needs to be a reason to compare on an equal basis the Main Wind Force Resisting System (MWFRS). However, the kicker was the height limits and the number of stories.

Thank you both (including Ben of course) for your help with this. I appreciate the assistance.

Dennis S. Wish, PE

California Professional Engineer

Structural Engineering Consultant

La Quinta, CA 92253

dennis.wish@verizon.net

http://structuralist.wordpress.com

http://www.structuralist.net

From: Acharya, Suresh [mailto:Suresh.Acharya@ci.concord.ca.us]
Sent: Thursday, July 31, 2008 12:49 PM
To: 'seaint@seaint.org'
Subject: RE: MWFRS Method 1 Question

Dennis,

Not 100% sure if you are referring to the simplified methods for wind for seismic, but there is (was) a mistake in Breyers' book in the wind force calculation example when wind is parallel to ridge. Did you notice it?

From: Dennis Wish [mailto:dennis.wish@verizon.net]
Sent: Thursday, July 31, 2008 12:31 PM
To: seaint@seaint.org
Subject: MWFRS Method 1 Question

Has ASCE 7-05 been revised to allow the use of wind design Method 1 (Simplified Design) for light-framed residential buildings with shear wall buildings to three story or is it still restricted to 2-story or 30-feet? Breyer's book on wood design makes a strong case for the use of the Simplified method but restricts the use of Method 1 to a two story shear wall building (wood frame). I thought the limit for both the MWFRS and LFRS were limited to 30-feet in height or three stories?

Dennis S. Wish, PE

California Professional Engineer

Structural Engineering Consultant

La Quinta, CA 92253

dennis.wish@verizon.net

http://structuralist.wordpress.com

http://www.structuralist.net

RE: MWFRS Method 1 Question

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RE: MWFRS Method 1 Question

Dennis,

From: Dennis Wish [mailto:dennis.wish@verizon.net]
Sent: Thursday, July 31, 2008 12:31 PM
To: seaint@seaint.org
Subject: MWFRS Method 1 Question

Dennis S. Wish, PE

California Professional Engineer

Structural Engineering Consultant

La Quinta, CA 92253

dennis.wish@verizon.net

http://structuralist.wordpress.com

http://www.structuralist.net

RE: MWFRS Method 1 Question

Method 1 applies to “low rise” buildings as defined on page 21. It could be up to 60 ft. high. No limitation on the number of stories.

Ben Yousefi, SE, CBO

Chief Building Official

City of Mountain View, CA

(650) 526-7007

ben.yousefi@mountainview.gov

From: Dennis Wish [mailto:dennis.wish@verizon.net]
Sent: Thursday, July 31, 2008 12:31 PM
To: seaint@seaint.org
Subject: MWFRS Method 1 Question

Has ASCE 7-05 been revised to allow the use of wind design Method 1 (Simplified Design) for light-framed residential buildings with shear wall buildings to three story or is it still restricted to 2-story or 30-feet? Breyer’s book on wood design makes a strong case for the use of the Simplified method but restricts the use of Method 1 to a two story shear wall building (wood frame). I thought the limit for both the MWFRS and LFRS were limited to 30-feet in height or three stories?

Dennis S. Wish, PE

California Professional Engineer

Structural Engineering Consultant

La Quinta, CA 92253

dennis.wish@verizon.net

http://structuralist.wordpress.com

http://www.structuralist.net

RE: ACI 318 App D, and wedge anchors

Jared Keyser wrote:

<< I have wondered about presenting the case that in wood construction,
if the anchors are sized such that the shear chord nailing fails prior
to developing the concrete breakout force in the anchor, you have
essentially provided a ductile yielding connected element. As such, you
have met the ACI provisions. This assumes that nail pullout is a
ductile failure mechanism, which in the case of cyclical seismic forces,
I believe this argument is valid, at least based on the high R-value
permitted to be used in design. The problem with this is QA/QC and
making sure the contractor does not over-nail the chord and force the
weak link back to the anchor.>>

This is the theory I use...it defaults you to D.3.3.5. For an 8"-wide
stemwall, You can make an HTT16 work w/ Simpson spec'd nailing, but not
an HTT22. I would not count on (nor encourage) a contractor to
under-nail an anchor.

The real problem, though, is that post-installed anchors work better, by
taking a bunch of slack out of the system. It's the only way I see the
anchor aligned w/ the shearwall chord. Often the cast in place anchors
are out by inches. Nobody wants to see the contractor torque the HD
strap over to the stud, although many do. So is the epoxy AB that
doesn't meet ACI 318 D.3.3.4 worse than the dynamic increase in load due
to poorly installed SSTBs? I don't think so, but it's a good question
you raise.

regards,
Gordon Goodell

MWFRS Method 1 Question

Has ASCE 7-05 been revised to allow the use of wind design Method 1 (Simplified Design) for light-framed residential buildings with shear wall buildings to three story or is it still restricted to 2-story or 30-feet? Breyer’s book on wood design makes a strong case for the use of the Simplified method but restricts the use of Method 1 to a two story shear wall building (wood frame). I thought the limit for both the MWFRS and LFRS were limited to 30-feet in height or three stories?

Dennis S. Wish, PE

California Professional Engineer

Structural Engineering Consultant

La Quinta, CA 92253

dennis.wish@verizon.net

http://structuralist.wordpress.com

http://www.structuralist.net

RE: ACI 318 App D, and wedge anchors

In one existing residential remodel job, I utilized a bracket that placed the post-installed mechanical anchors in shear on the face of the foundation wall and was able to meet the requirements in a retrofit situation.

I have wondered about presenting the case that in wood construction, if the anchors are sized such that the shear chord nailing fails prior to developing the concrete breakout force in the anchor, you have essentially provided a ductile yielding connected element. As such, you have met the ACI provisions. This assumes that nail pullout is a ductile failure mechanism, which in the case of cyclical seismic forces, I believe this argument is valid, at least based on the high R-value permitted to be used in design. The problem with this is QA/QC and making sure the contractor does not over-nail the chord and force the weak link back to the anchor.

Jared

From: Rhkratzse@aol.com [mailto:Rhkratzse@aol.com]
Sent: Thursday, July 31, 2008 10:01 AM
To: Jared Keyser; Seaint@seaint.org
Subject: Re: ACI 318 App D, and wedge anchors

In a message dated 7/31/08 10:02:24 AM, jkeyser@lcmf.com writes:

or avoid post-installed anchors entirely if possible.

In other words, completely *prohibit* the improvement of the earthquake strength and safety of existing structures, except with (often prohibitively expensive) new foundation construction.

I'm thinking mainly of single-family homes, where many owners are willing to invest a "reasonable" amount in anchor bolts and plywood, but not an order of magnitude more in a whole new foundation.

Ralph Hueston Kratz, S.E.
Richmond CA USA

**************
Get fantasy football with free live scoring. Sign up for FanHouse Fantasy Football today.
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re: EQ and international codes

OK, not completely a technical post.. Sorry, delete now if you want!

Harold had an interesting point about economics dictating a country or region's quality of construction, adherence to codes, and engineering/architectural knowledge. This would make a great ASCE/Structure magazine article. California has lead the world in seismic engineering, design and codes, and is also an economic powerhouse that would rank it between 7-10 of the top economies of the rest of the world (depending on who you ask). Chicago, NYC and Boston have also been forerunners in engineering and construction due to their economies and therefore the need to go vertical with buildings and big with bridges.

You could say the same for economics, need, and knowledge in countries such as Japan (bridges & buildings), France (buildings, dams and viaducts), Netherlands (water and land management), Scandanvia with ships and oil rigs, Switzerland and Austria with viaducts and tunnels, etc etc... And the list goes on. If you have money as a country or region, and you need complex buildings and infrastructure, you can develop them and maintain them, and develop the knowledge and skill base you need to accomplish these things.

From what I understand about India is there are lots of university trained engineers and scientists that cannot find work in their economy. They certainly have a knowledge base capable of better construction and engineering, but if there simply is no money for it, it cannot be done. I would imagine the same for China. Other things contributing I am sure are the lack of available materials like Harold said, proper equipment, all of the standards and testing agencies, and local cultural issues such as nepotism, cronyism, bribery, etc. And in China who knows what happens in a communist country, if someone wants something built fast and cheap, they can probably steamroll it through. There are not lawyers on every corner either, and a court system that will support legal claims....

I noticed in two developing countries that I travelled to, Peru and Thailand, that in the big cities of Lima and Bangkok, you see big concrete highrises as modern as anything we have in the US. The skytrain in Bangkok is as modern as any city, and the toll road out of Lima makes you think you are at home... Now you go out into the country, and in Peru you see mud brick factories- somebody with a small plot of land, some mud, and a kiln... And that is what they build their houses out of. In Thailand you go out into the country and there are houses on rivers on bamboo poles. You make due with what you have, better to be dry and protected in a building that does not meet code then to be wet and miserable outside!

You see this in this country also, ever travel around the rural south? My ex wife from Spain was amazed by our manufactured and mobile homes, it was shocking to her that in the US we would have this type of construction.

To rip off Churchhill, "The US has the worse building code system in the world, until you compare it to all of the others..."

Andrew Kester, PE
Principal/Project Manager
ADK Structural Engineering, PLLC
1510 E. Colonial Ave., Suite 301
Orlando, FL 32803

RE: ACI 318 App D, and wedge anchors

Jared -

The ductility issue (1/2.5) not only applies to tension anchors but shear
anchors as well.

Just curious, what "series of calcs" did you go through? Section D.3.3.5
isn't clear on this.

T. William (Bill) Allen, S.E.
ALLEN DESIGNS
Consulting Structural Engineers

> -----Original Message-----
> From: Jared Keyser [mailto:jkeyser@lcmf.com]
> Sent: Thursday, July 31, 2008 9:54 AM
> To: Seaint@seaint.org
> Subject: RE: ACI 318 App D, and wedge anchors
>
> I'm curious on if you all are meeting the ductile failure requirements in
> high seismic zones due to Section D.3.3.4 . I went through a series of
> calcs with the Hilti products and found that it was not possible to make
> the ductile steel elements fail with the current manufactured anchors. I
> emailed Hilti tech support and they confirmed this is the case. As such,
> if you have tension in your post installed anchors in high seismic zones,
> you are limited to providing either 2.5 times the anchors in a non-ductile
> failure mode, provide a link in the connecting element that will fail in a
> ductile manner before the anchor pulls out, or avoid post-installed
> anchors entirely if possible.
>
> Jared Keyser
>
>
>
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plywood diaphragm to light gage

Several years ago I believe I ask a question of the list to see if there was any diaphragm values for plywood screwed to light gage supports. If I remember correctly, someone posted some ICBO testing (maybe ICC is newer) that contained diaphragm values. It may have been shot pins, I just don’t remember. I have searched through my library, but since my office has moved a couple of times since then I can’t find that information (if it existed in the first place). If anyone out there has that information, I would appreciate a copy or a link to the information. I have a client that is considering a roof system that would require this, if possible.

Thanks,

Joe Grill

Joseph R. Grill, PE

Verde Valley Engineering, PLLC

2220 Sky Drive

Clarkdale, AZ 86324

Ph. (928) 600-5459

Fax (928) 649-3659

email: VVEng@cableone.net

Re: Staad 2007

On Jul 31, 2008, at 9:20 AM, Scott Maxwell wrote:

> Many software companies are going or have gone to a subscription
> based update based model rather than the traditional "pay for major
> upgrade" based models. And it is NOT just companies that do
> structural software (although, I see it on a much more regular
> basis for more "specialized" software, such as structural software
> and CADD software, that would be used in design offices). Even
> Microsoft has talked about doing it (they have not had the guts to
> pull the trigger on it yet) for programs like Office.

This kind of thing is characteristic of 'vertical market' software
and it's the response to the disappearance of the LAGS-driven user
who is obsessed with having the newest hardware and software. It's a
big advantage to the developer because the customer is held captive.
Every user pays for every change, even if the change is buggy or
irrelevant to the customer's effort. ANSYS Inc has been using this
model for years and they've been making scads of money on it, but the
overall usability hasn't improved much and bugs keep showing up.
COSMOS is doing the same thing, and I understand AutoCAD and several
other CAD packages are doing it. It seems like part of the MicroSoft
business model where nothing is backwards compatible and new versions
of applications usually mean new hardware and (lucky for MS) a new OS.

So yeah. I'm a cynic. Hardware is getting better but software is
getting worse and more expensive to boot. The 'subscription-based'
model is just the latest gimmick for squeezing more money out of a
product without actually making it better. Fortunately for me Apple
has succumbed to a much smaller degree, so I can run pretty much any
software written in the last 15 years on my networked mix of laptops,
old OS 9 machinery and OS X computers. God help me if I were a
Windows user--I'd be in Dennis Wish's shoes of reverting to manual
calculations--not real easy when you're doing structural dynamics.

Christopher Wright P.E. |"They couldn't hit an elephant at
chrisw@skypoint.com | this distance" (last words of Gen.
.......................................| John Sedgwick, Spotsylvania
1864)
http://www.skypoint.com/~chrisw/

Re: ACI 318 App D, and wedge anchors

In a message dated 7/31/08 10:02:24 AM, jkeyser@lcmf.com writes:

or avoid post-installed anchors entirely if possible.

RE: Mexico Building Code

Thank you Ben & Tom

Rich

From: Ing. Benjamín Arcos Reyes [mailto:barcosr@gmail.com]
Sent: Thursday, July 31, 2008 10:54 AM
To: seaint@seaint.org
Subject: Re: Mexico Building Code

Current building code in Mexico was published in 2004 with some important changes respect to former editions. The main part is the Building Code for Federal District (Reglamento de Construcciones para el Distrito Federal) with general regulations and it contains a series of additional publications called Complementary Technical Regulations for Design and construction, which include the design criterion and specifications for concrete, steel, masonry, foundations, wood, seismic design, wind design, loads for design and architectural projects. Those regulations and the code were specified for desing of structures in the Federal District, anyway, with the exception of seismic and wind design are commonly used everywhere in Mexico.

In order to determine wind and seismic loads it is very helpful the Civil works Design Handbook (Manual de Diseño de Obras Civiles) published by the Electricity Federal Comision (CFE) because it covers almost all the country.

Main cities have their own building code, but many of them are based on the Mexico city building code.

Design regulations of Mexico City building code are based on LRFD. There is also a handbook edited by Steel Construction Mexican Institute (Instituto Mexicano de Construcción en Acero) which is based on ASD and is also commonly used. Many structural firms also prefer to use the AISC Specification, both in ASD and for LRFD.

On Wed, Jul 30, 2008 at 8:31 PM, Rich Lewis <seaint04@lewisengineering.com> wrote:

I did a search of the archives and couldn't find an answer. Sorry if this topic has been dealt with before.

What is the building code of Mexico? Have they adopted an ICC code? Do they have a national code, or do states or cities have the option to adopt different codes?

Thanks.

Rich

--
Atentamente,
\0>
.||
< \_

Ing. Benjamín Arcos Reyes.
Visita: http://www.senderodelpeje.blogspot.com/