Structural Engineering: 2009-01-18

Saturday, January 24, 2009

Re: White Male Professionals

Here's a crazy idea. How's about the gov't, who created this mess in
the first place, stop deciding who is "worthy" to receive stolen
money? Cut taxes and spending to the bone and allow the People to take
care of themselves

William L. Polhemus, Jr. P.E.
Via iPhone 3G

On Jan 23, 2009, at 8:05 PM, "Rich Lewis"
<seaint04@lewisengineering.com> wrote:

> I'm not sure where he's been lately. Here in Texas the majority of
> construction people are not white males.
>
>
>
> -----Original Message-----
> From: Michel Blangy [mailto:mblangy@satco-inc.com]
> Sent: Friday, January 23, 2009 10:34 AM
> To: Seaint@Seaint. Org
> Subject: White Male Professionals
>
> Economist and former U.S. Labor Secretary Robert Reich worries about
> too
> much of the Obama stimulus going to white males in the construction
> sector.
>
> "If there aren't enough skilled professionals to do the jobs
> involving new
> technologies, the stimulus will just increase the wages of the
> professionals
> who already have the right skills rather than generate many new jobs
> in
> these fields. And if construction jobs go mainly to white males who
> already
> dominate the construction trades, many people who need jobs the most
> --
> women, minorities, and the poor and long-term unemployed -- will be
> shut
> out," Reich said.
>
> Speechless.
>
> ----------------------
> Michel Blangy, P.E.
> White Male Professional
>
>
>
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Friday, January 23, 2009

RE: Fire Wall Connection

I have had the nylatron approach accepted by building officials in a number of different states.

________________________________

From: Rich Lewis [mailto:seaint04@lewisengineering.com]
Sent: Fri 1/23/2009 8:57 PM
To: seaint@seaint.org
Subject: RE: Fire Wall Connection

Stuart,

Thank you for this information. I've thought about a melt-away type bolt,
but I wasn't sure this was an acceptable solution. What if the fire is
sufficiently far enough away not to cause the melt temperature, yet a
collapse pulls on the steel due to a collapse in the next bay or so? Is the
thinking that the fire was hot enough to cause a collapse nearby, it must be
hot enough at the bolts in order to melt it? Is this a common rational?
Has anyone published a paper on this that addresses wall bracing concepts?

Thanks again for your help.

Rich

From: Stuart, Matthew [mailto:mStuart@cmxengineering.com]
Sent: Friday, January 23, 2009 11:36 AM
To: seaint@seaint.org
Subject: RE: Fire Wall Connection

Connect the wall to each structure using Nylatron bolts. A ceramic/plastic
composite material that "melts" at 500 degrees, or almost 1/2 the
temperature that steel starts to deform. The material has high shear
capacity and is ideal for collector force connection to the wall.

_____

From: seaint04@lewisengineering.com [mailto:seaint04@lewisengineering.com]
Sent: Fri 1/23/2009 11:52 AM
To: Seaint
Subject: Fire Wall Connection

I know this has been addressed in previous messages. Unfortunately I'm
having trouble searching the SEAInt archives. Every time I put in a
search, no matter which radio button I push, it does a general web search
and not an exclusive archive search.

I have a 2 story steel framed building with a masonry fire separation wall.
IBC 702 definitions requires the wall to stand if the structure on either
side of the wall collapses. Ideally I would like to use the fire wall as a
shear wall for one side of the wall. I realize this may not be practical.
I'm wondering if someone has developed a detail for attaching to the wall
as a shear wall but breaking away when needed for a fire event.

I have seen someone else use a detail where they bolted an angle to the
face of the wall and welded a smooth stud to the bottom of the outstanding
leg. They then welded a horizontal plate from the steel beam that had a
hole in it for the stud to slide up and down. If the beam went down far
enough the stud pulled out of the hole. The idea is interesting, but I
think it has a flaw. If the beam is pulled horizontal, say towards the
center of collapse, it still pulls horizontally on the wall. I thought
maybe using a slot perpendicular to the wall instead of a hole, to let the
beam pull away, yet have resistance for a shear wall force, but I guess I
can't guarantee the beam would only be pulled in that direction due to a
collapse.

Can anyone describe to me a better detail?

Thanks for your help.

Rich

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RE: White Male Professionals

I'm not sure where he's been lately. Here in Texas the majority of
construction people are not white males.

-----Original Message-----
From: Michel Blangy [mailto:mblangy@satco-inc.com]
Sent: Friday, January 23, 2009 10:34 AM
To: Seaint@Seaint. Org
Subject: White Male Professionals

Economist and former U.S. Labor Secretary Robert Reich worries about too
much of the Obama stimulus going to white males in the construction sector.

"If there aren't enough skilled professionals to do the jobs involving new
technologies, the stimulus will just increase the wages of the professionals
who already have the right skills rather than generate many new jobs in
these fields. And if construction jobs go mainly to white males who already
dominate the construction trades, many people who need jobs the most --
women, minorities, and the poor and long-term unemployed -- will be shut
out," Reich said.

Speechless.

----------------------
Michel Blangy, P.E.
White Male Professional

RE: Visual Analysis for diaphragms

Gordon,

I suggest writing Terry Kubat at tech support. He has always responded to my questions promptly with good information. There email is support@iesweb.com

Rich

From: Gordon Goodell [mailto:GordonGoodell@harmonydesigninc.com]
Sent: Friday, January 23, 2009 8:17 AM
To: seaint@seaint.org
Subject: Visual Analysis for diaphragms

For those of you out there using Visual Analysis, has anyone ever tried to model semi-rigid or rigid diaphragm behaviour in VA? Probably not worth it for simple configurations, but for arrangements w/ horiz/vert irregularities it seems like it could be a useful analysis tool.

thanks,

Gordon Goodell

RE: Fire Wall Connection

Stuart,

Thanks again for your help.

Rich

From: Stuart, Matthew [mailto:mStuart@cmxengineering.com]
Sent: Friday, January 23, 2009 11:36 AM
To: seaint@seaint.org
Subject: RE: Fire Wall Connection

_____

From: seaint04@lewisengineering.com [mailto:seaint04@lewisengineering.com]
Sent: Fri 1/23/2009 11:52 AM
To: Seaint
Subject: Fire Wall Connection

Can anyone describe to me a better detail?

Thanks for your help.

Rich

******* ****** ******* ******** ******* ******* ******* ***
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Re: SIPS in Seismic Design Category D, E ,F

Short answer, no.

Longer answer, depends on what you define as acceptance. To my knowledge,
there are no SIP manufacturers that have an ICC-ES Evaluation report that
evaluates SIPs for use in SDC D, E, or F per the 2006 IBC. This is because
the ICC-ES has _NO_ Acceptance Criteria for evaluating SIPs for use in SDC
D, E, or F per the IBC 2006. They did approve an Appendix A for AC 04 a
couple of years ago that would have allowed an evaluation method for SIPs
as shearwalls in SDC D, E, and F, but the withdrew it about a year or so
ago. As a result, the only evaluation reports that I know of for "high
seismic" areas are legacy reports that evaluated under the 1997 UBC and
maybe the 2000 IBC.

Now, that does NOT mean you cannot use SIPs. It just means that you need
to talk manufacturers about what information that you can get from them
that you can then use to try to get the local code official to accept the
use of SIPs on a case by case basis...keep in mind that an ICC-ES report is
NOT in and of itself "acceptance"...strictly speaking it is PURELY an
evaluation report and your local code official is still the one that muse
officially accept the use of the item evaluated...it just turns out that
most code officials will automatically accept anything that has an ICC-ES
report.

If you are considering using Insulspan SIPs, then contact me offline and I
can get you a technical bulletin covering seismic use that you can see if
the code official will accept. If you are using some other manufacturer,
then you will have to contact them to see if they have a technical report
on seismic testing that they can provide to you for you to try to use with
the local code official. I know that Premier has done some testing with
the Appendix A before it was withdrawn, but I don't know if they have a
technical bulletin or report that they have prepared or not. You can
certainly call them and asked them if they have anything. And I would say
that there is a decent chance that R-Control might have something as well.
If you are looking at smaller SIP manufacturers, you might be out of luck.
I am only aware of the "big three" having potentially proceeded with
Appendix A testing (i.e. seismic testing) prior to it being withdrawn.

In addition to that, there was some "generic" testing done by APA as part
of the effort to add SIPs to the 2007 Supplement to the IRC. They did some
limited shearwall testing...I believe, if my memory serves me correct, some
of it was dynamic shearwall tests. Don't know if there is enough of report
that could be used to convince a code official or not. In addition, just
this past year, some testing was completed at Penn State. I am pretty sure
it was dynamic shearwall testing. Don't know if they have any published
reports or not. The testing was sponsored by SIPA, so if you want to see
if there is anything publicly available, you could try contact SIPA
(www.sips.org). This testing was mainly to test the different fastener and
spline types, but it potentially could be used to try to get code official
approval...maybe.

Feel free to contact me offline and we could schedule a time to talk by
phone if you want. My schedule is a little screwed up due to medical
issues, but I am available if you need some "consultation".

Regards,

Scott
Adrian, MI

On Fri, 23 Jan 2009 13:42:49 -0800, "Gerard Madden, SE"
<gmse4603@gmail.com> wrote:
> Anyone out there (Scott Maxwell, white courtesy phone) know of the
> acceptance of SIPS in SDC D or worse in California?
>
> City of LA, DSA etc... are those guys willing to consider them?
>
> tia,
> -gm
>

Re: SIPS in Seismic Design Category D, E ,F

I found something to justify them in plan review, but sorry cannot recall. Try some manufacturers

>>> "Gerard Madden, SE" <gmse4603@gmail.com> 01/23/2009 3:09 PM >>>
Not last time I checked (4 years ago)...

-gm

On Fri, Jan 23, 2009 at 2:23 PM, Paul Guthrie <PGuthrie@simivalley.org> wrote:

Got an ICC #?

Paul Guthrie, PE
Building & Safety
City of Simi Valley
805.583.6885

>>> "Gerard Madden, SE" <gmse4603@gmail.com> 01/23/2009 1:42 PM >>>

Anyone out there (Scott Maxwell, white courtesy phone) know of the acceptance of SIPS in SDC D or worse in California?

City of LA, DSA etc... are those guys willing to consider them?

tia,
-gm

RE: SIPS in Seismic Design Category D, E ,F

Try: ESR-1138

Not sure about SDC D.

From: Gerard Madden, SE [mailto:gmse4603@gmail.com]
Sent: Friday, January 23, 2009 1:43 PM
To: seaint@seaint.org
Subject: SIPS in Seismic Design Category D, E ,F

Anyone out there (Scott Maxwell, white courtesy phone) know of the acceptance of SIPS in SDC D or worse in California?

City of LA, DSA etc... are those guys willing to consider them?

tia,
-gm

Re: SIPS in Seismic Design Category D, E ,F

Not last time I checked (4 years ago)...

-gm

On Fri, Jan 23, 2009 at 2:23 PM, Paul Guthrie <PGuthrie@simivalley.org> wrote:

Got an ICC #?

Paul Guthrie, PE
Building & Safety
City of Simi Valley
805.583.6885

>>> "Gerard Madden, SE" <gmse4603@gmail.com> 01/23/2009 1:42 PM >>>

Anyone out there (Scott Maxwell, white courtesy phone) know of the acceptance of SIPS in SDC D or worse in California?

City of LA, DSA etc... are those guys willing to consider them?

tia,
-gm

Re: SIPS in Seismic Design Category D, E ,F

Got an ICC #?

Paul Guthrie, PE
Building & Safety
City of Simi Valley
805.583.6885

>>> "Gerard Madden, SE" <gmse4603@gmail.com> 01/23/2009 1:42 PM >>>
Anyone out there (Scott Maxwell, white courtesy phone) know of the acceptance of SIPS in SDC D or worse in California?

City of LA, DSA etc... are those guys willing to consider them?

tia,
-gm

SIPS in Seismic Design Category D, E ,F

Anyone out there (Scott Maxwell, white courtesy phone) know of the acceptance of SIPS in SDC D or worse in California?

City of LA, DSA etc... are those guys willing to consider them?

tia,
-gm

RE: Fire Wall Connection

Connect the wall to each structure using Nylatron bolts. A ceramic/plastic composite material that "melts" at 500 degrees, or almost 1/2 the temperature that steel starts to deform. The material has high shear capacity and is ideal for collector force connection to the wall.

________________________________

From: seaint04@lewisengineering.com [mailto:seaint04@lewisengineering.com]
Sent: Fri 1/23/2009 11:52 AM
To: Seaint
Subject: Fire Wall Connection

Can anyone describe to me a better detail?

Thanks for your help.

Rich

******* ****** ******* ******** ******* ******* ******* ***
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Fire Wall Connection

Can anyone describe to me a better detail?

Thanks for your help.

Rich

White Male Professionals

Economist and former U.S. Labor Secretary Robert Reich worries about too
much of the Obama stimulus going to white males in the construction sector.

Speechless.

----------------------
Michel Blangy, P.E.
White Male Professional

RE: Smart Columns Article

The article calls these "Smart steel columns". I guess they know better than to fracture at the wrong time. Sort of a Chuck Norris thing.

Seriously, I read the article but it's going to take more time than I have right now for me to study his ideas.

Bob Garner

From: Jeremy White [mailto:jwhite@megr.com]
Sent: Friday, January 23, 2009 5:36 AM
To: seaint@seaint.org
Subject: Smart Columns Article

Did anyone read the "Designing for horizontal movement" article in the Dec '08 issue of Structural Engineer magazine? The phrase "again reach full plastification" raised a flag for me. Are these columns intended to work by exceeding there elastic stress, and doing it more than once? Am I missing something?

- Jeremy

Visual Analysis for diaphragms

thanks,

Gordon Goodell

Smart Columns Article

- Jeremy

Thursday, January 22, 2009

RE: Lateral Stability of a Box Beam ?

Mr T. Boene,

Bills request for information for design of a box beam was to replace two
independent beams, with a single box beam, because if the effective length
is the full span, the two independent beams get too large. The choice of the
box beam was indicated to be to increase torsional resistance. There was no
mention of providing a simple plywood diaphragm to brace the two beams, or
any other form of bracing. Turning two beams into a box beam potentially
wider than deep, instead of providing bracing is not very efficient.

And yes the Australian timber structures code (AS1720) provides for what it
calls spaced columns: these can have either packer plates or batten plates
joining the sticks. It also covers built up sections: I-beams, boxes. The
steel structures code (AS4100) covers laced and battened columns.

Whether the two beams are considered as an assembly forming a horizontal
truss, or a battened column. Bracing the two beams would seem to be more
efficient than adding the extra weight of plywood covers top and bottom for
the full length, to lightly loaded beams: not supporting much more than
their own self weight.

Whilst Paul may be correct about the two beams displacing in unison:
implying a half sine wave buckle for the full span. The battened column
analogy indicates that such will be at a higher axial load, than for the
beams acting individually.

Galambos discusses lean on effects for columns and beams: and beams don't
appear to be a major concern (pg 471?). More over then goes on to design
bracing, which is similar to the bridging we use in Australia for coldformed
girts/purlins.

Which is where I think the issue is. Most of the photos I have seen of
bridging of purlins in the US, show bracing which appears little better than
sag rods. So my alternative proposal of providing bridging to the beams
wasn't a good one. Divided by a common language.

And you are right we are not disputing the benefits of box beams. Rather
whether a box beam is more efficient or suitable for the situation than
bracing. If bracing is the alternative, then suitable bracing then becomes
an issue.

Bill no doubt will accuse me of over complicating things again, for what is
probably a simple design problem.

Regards
Conrad Harrison
B.Tech (mfg & mech), MIIE, gradTIEAust
mailto:sch.tectonic@bigpond.com
Adelaide
South Australia

Re: Lateral Stability of a Box Beam ?

Mr Ransom,

What are simple ties by the way? Did you mean pin-pin connection for the bracing members? If so, why not consider the "ladder" as a parallel-chord truss? Do not truss transfer shear? Plywooded trusses are even better -- consider them as a horizontal shear wall.

Can compare with the laced and battened columns and struts that we see in old steel bridges. British steel code used to have a detailed method of designing laced and battened members to carry axial and bending. How about Australian code, Mr. Conrad?

Bill was asking for a method of designed boxed beams. Nobody here is arguing that boxed beams has no benefit. I am not sure what point you are trying to make here.

> -----Original Message-----
> From: Paul Ransom [mailto:ad026@hwcn.org]
> Sent: Sunday, 18 January 2009 06:54
> To: seaint@seaint.org
> Subject: Re: Lateral Stability of a Box Beam ?

> If the 2 beams are identical with identical loading, there is no benefit to
> simple ties. Unless there is some shear resistance mechanism between the
> brace points, the unbraced length for lateral buckling is not reduced and
> the ladder simply behaves like 2 tied beams (e.g. Iy1 + Iy2) and they just
> displace in unison.

> Regards
> Paul
> --
> Paul Ransom, P.Eng.
> ph 905 639-9628
> fax 905 639-3866
> ad026@hwcn.org

Wednesday, January 21, 2009

RE: Lateral Stability of a Box Beam ? (Bracing two independent beams by turning into single box beam)

Thanks Paul,

As you imply the published literature has extensive coverage of elastic
instability.

Not forgetting the original problem: two lightly loaded independent beams
with a concern that without lateral bracing the beams would be excessive in
size and possibly weight. The proposed solution was, given the two beams
were in close proximity, to turn them into a single box section, to increase
torsional resistance and therefore improve lateral stability. If the two
independent beams were further apart this would not be a very practical or
efficient solution.

Considering the box section. It principally consists of two pairs of
flanges. One pair to resist bending in the vertical plane and the other pair
to resist bending in the horizontal plane. Since the objective is to
constrain the size of the two beams, expect the box to be wider than it is
deep. Not very efficient to support the principal vertical loading. For
roughing out the size of the box the elastic section modulus Z for a pair of
flanges is approximately Z=DBT(approx). This is obtained by ignoring the
relatively insignificant contribution of Iy1+Iy2 when applying the parallel
axis theorem. D=depth between flanges, B=breadth, T=thickness. The two
proposed plywood covers, would provide the flanges for the vertical loading,
making the two original beams somewhat redundant. Being relatively broad
flanges and thin, the plywood may be prone to local buckling. In the
horizontal plane, the two proposed beams provide the flanges for a
relatively much deeper beam, with relatively more stable flanges. This deep
beam resists the pseudo horizontal displacing force, causing the buckling.
So potentially being much stronger in the horizontal plane than the vertical
plane, the lateral buckling problem has been reduced. But not very
efficiently.

Staying with the beam analogy for the assembly of the two beams. To provide
resistance to the lateral displacement: there is more than one way to create
such a beam in the horizontal plane. For example the two beams can be
considered the chords of a parallel chord triangulated truss, or a
Vierendeel girder. Offcuts from the section used for the two beams can be
used as perpendicular webs. These fastened top and bottom of the depth of
the beam to provide some torsional resistance to minimise twisting of the
two beams which are otherwise simple plate elements. Then continuing with
the beam/truss analogy, provide steel strap cross-bracing for the diagonal
webs (shear). Crossed because the pseudo displacing force could displace or
load in either direction.

However, returning to the original problem: it is not about forming a beam
in the horizontal plane: it is about changing the mode of buckling.
Independently the two beams (plates) will buckle their full span in a half
sine wave. If an offcut of the beam section is fastened at midspan: then for
displacement to occur at midspan either the offcut/bridge is placed in
tension as the beams displace away from each other, or the offcut/bridge is
placed in compression and buckles or crushes, as the beams displace towards
each other. With the bridge designed to have the appropriate stiffness, then
each individual beam/plate now buckles with a full sine wave, or a half sine
wave in half its length.

If not comfortable with such. Then as I first indicated calculate the axial
compression force causing the buckling, and treat the assembly as either
spaced columns, battened column or laced column. The differences are
described by Galambos. If adopt the laced column then effectively returning
to forming a truss in the horizontal plane.

Since the bending moment typically varies along the length of the beam, the
normal compressive forces/stresses on the section also vary. So primarily
trying to reduce the segment length of the most highly stressed portion of
the beam. So not entirely necessary to box up the entire span. And if
plywood is the preference then diaphragm bracing to one side may be more
practical then boxing.

After all are the plywood covers going to be installed in situ, whilst the
two beams are propped? If so how easy is it to install the lower plywood
cover? If the box is fabricated first, then may have transportation and
crane problems. For that matter are timbers 27ft long readily available?

Since the beams are lightly loaded: not much more than self-weight. Then can
probably load test during construction, and add additional restraint as
necessary to push member bending capacity closer to section bending
capacity.

But each to their own comfort zone.

Regards
Conrad Harrison
B.Tech (mfg & mech), MIIE, gradTIEAust
mailto:sch.tectonic@bigpond.com
Adelaide
South Australia

RE: Bond Strength of Portland Grouts

This is correct. In the old ACI 318-63, there was a bond strength for plain bar. This proved to be HIGHLY variable to zero bond with some smooth cold rolled bars. Stick with the tested information.

Regards, Harold Sprague

To: seaint@seaint.org
Subject: Re: Bond Strength of Portland Grouts
From: Tom.Hunt@fluor.com
Date: Wed, 21 Jan 2009 15:22:58 -0800

Thor,

It is a lot more complicated than simple "bond" strength. The "bond" strength of your grout to the rod will be different depending on whether you are using smooth bar, threaded rods, or deformed rebar and the different values you see for each in a HILTI catalog are based on testing. If my memory is correct, in the "good old days", we use to use 60 psi actual bond of concrete to steel rods but I suspect it would be much higher for a treaded rod. If your client does not like the epoxy suggest a proprietary cementitious grout which again would have an ICC report with values you can fall back on otherwise you are left holding the bag if anything goes wrong. You would also have to come up with your own written installation, curing, testing, and inspection guidelines instead of being able to just reference an appropriate ICC ES Report number.

Thomas Hunt, S.E.
Fluor

"Thor Tandy" <vicpeng@telus.net>
01/21/2009 02:36 PM
Please respond to seaint

To	"SEAINT" <seaint@seaint.org>
cc
Subject	Bond Strength of Portland Grouts

I've been asked by the Client to review substituting Hilti RE-500 SD epoxy with "Portland Expanding (no-shrink) Grout". I'm anchoring a cable support post, using baseplate and threaded ASTM 193 rods, to conglomerate bedrock To do that I need to have an idea of the available bond strength. The supplier I was able to contact said that it isn't normal to have BS for grouts. I don't think it's doable but before I cause a frown on my client's face, does anyone have a reasonable value for the bond strength of this grout. The RE-500-SD has up to 12.4MPa (1800psi) but I don't think the grout is anywhere near this. TIA Thor A. Tandy P.Eng, C.Eng, Struct.Eng, MIStructE Victoria, BC Canada ******* ****** ******* ******** ******* ******* ******* *** * Read list FAQ at: http://www.seaint.org/list_FAQ.asp * * This email was sent to you via Structural Engineers * Association of Southern California (SEAOSC) server. To * subscribe (no fee) or UnSubscribe, please go to: * * http://www.seaint.org/sealist1.asp * * Questions to seaint-ad@seaint.org. Remember, any email you * send to the list is public domain and may be re-posted * without your permission. Make sure you visit our web * site at: http://www.seaint.org ******* ****** ****** ****** ******* ****** ****** ********

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RE: Bond Strength of Portland Grouts

Thanks Tom.

Yes, I agree. I just wanted to compare apples with apples. With threaded rod per Hilti BS = 12.4 MPa. 60 psi gives about 0.4MPa, which is close to that suggested by Target Products Ltd.

I've told the client to stay with the epoxy.

Thor

-----Original Message-----
From: Tom.Hunt@fluor.com [mailto:Tom.Hunt@fluor.com]
Sent: Wednesday, January 21, 2009 3:23 PM
To: seaint@seaint.org
Subject: Re: Bond Strength of Portland Grouts

Thor,

It is a lot more complicated than simple "bond" strength. The "bond" strength of your grout to the rod will be different depending on whether you are using smooth bar, threaded rods, or deformed rebar and the different values you see for each in a HILTI catalog are based on testing. If my memory is correct, in the "good old days", we use to use 60 psi actual bond of concrete to steel rods but I suspect it would be much higher for a treaded rod. If your client does not like the epoxy suggest a proprietary cementitious grout which again would have an ICC report with values you can fall back on otherwise you are left holding the bag if anything goes wrong. You would also have to come up with your own written installation, curing, testing, and inspection guidelines instead of being able to just reference an appropriate ICC ES Report number.

Thomas Hunt, S.E.
Fluor

"Thor Tandy" <vicpeng@telus.net>
01/21/2009 02:36 PM
Please respond to seaint

To
"SEAINT" <seaint@seaint.org>

cc

Subject
Bond Strength of Portland Grouts

I've been asked by the Client to review substituting Hilti RE-500 SD epoxy with "Portland Expanding (no-shrink) Grout". I'm anchoring a cable support post, using baseplate and threaded ASTM 193 rods, to conglomerate bedrock To do that I need to have an idea of the available bond strength. The supplier I was able to contact said that it isn't normal to have BS for grouts. I don't think it's doable but before I cause a frown on my client's face, does anyone have a reasonable value for the bond strength of this grout. The RE-500-SD has up to 12.4MPa (1800psi) but I don't think the grout is anywhere near this. TIA Thor A. Tandy P.Eng, C.Eng, Struct.Eng, MIStructE Victoria, BC Canada ******* ****** ******* ******** ******* ******* ******* *** * Read list FAQ at: http://www.seaint.org/list_FAQ.asp * * This email was sent to you via Structural Engineers * Association of Southern California (SEAOSC) server. To * subscribe (no fee) or UnSubscribe, please go to: * * http://www.seaint.org/sealist1.asp * * Questions to seaint-ad@seaint.org. Remember, any email you * send to the list is public domain and may be re-posted * without your permission. Make sure you visit our web * site at: http://www.seaint.org ******* ****** ****** ****** ******* ****** ****** ********
------------------------------------------------------------ The information transmitted is intended only for the person  or entity to which it is addressed and may contain  proprietary, business-confidential and/or privileged material.   If you are not the intended recipient of this message you are  hereby notified that any use, review, retransmission, dissemination,  distribution, reproduction or any action taken in reliance upon  this message is prohibited. If you received this in error, please  contact the sender and delete the material from any computer.    Any views expressed in this message are those of the individual  sender and may not necessarily reflect the views of the company.   ------------------------------------------------------------ 

Re: Bond Strength of Portland Grouts

Thor,

It is a lot more complicated than simple "bond" strength. The "bond" strength of your grout to the rod will be different depending on whether you are using smooth bar, threaded rods, or deformed rebar and the different values you see for each in a HILTI catalog are based on testing. If my memory is correct, in the "good old days", we use to use 60 psi actual bond of concrete to steel rods but I suspect it would be much higher for a treaded rod. If your client does not like the epoxy suggest a proprietary cementitious grout which again would have an ICC report with values you can fall back on otherwise you are left holding the bag if anything goes wrong. You would also have to come up with your own written installation, curing, testing, and inspection guidelines instead of being able to just reference an appropriate ICC ES Report number.

Thomas Hunt, S.E.
Fluor

"Thor Tandy" <vicpeng@telus.net>
01/21/2009 02:36 PM
Please respond to seaint

To	"SEAINT" <seaint@seaint.org>
cc
Subject	Bond Strength of Portland Grouts

------------------------------------------------------------ The information transmitted is intended only for the person  or entity to which it is addressed and may contain  proprietary, business-confidential and/or privileged material.   If you are not the intended recipient of this message you are  hereby notified that any use, review, retransmission, dissemination,  distribution, reproduction or any action taken in reliance upon  this message is prohibited. If you received this in error, please  contact the sender and delete the material from any computer.    Any views expressed in this message are those of the individual  sender and may not necessarily reflect the views of the company.   ------------------------------------------------------------

Bond Strength of Portland Grouts

To do that I need to have an idea of the available bond strength. The
supplier I was able to contact said that it isn't normal to have BS for
grouts. I don't think it's doable but before I cause a frown on my client's
face, does anyone have a reasonable value for the bond strength of this
grout.

The RE-500-SD has up to 12.4MPa (1800psi) but I don't think the grout is
anywhere near this.

TIA

Thor A. Tandy P.Eng, C.Eng, Struct.Eng, MIStructE
Victoria, BC
Canada

RE: Availability of Older AutoCAD Versions

List:

This is true... although not the original version.

Price is even cheaper when you visit the country it originates from.

Thanks,

Julius

Engr. Julius Micayas
P.E. license no.32969
Project Manager/Sr Lead Structural Engineer
River Consulting LLC.
#5 Sanctuary Blvd., Suite 101
Mandeville, Louisiana 70471
Phone - 985-624-1314 (direct)
985 624-1300 (office)
Fax - 985-624-1399
E-mail: jmicayas@riverconsulting.com
W-page: www.riverconsulting.com and www.kindermorgan.com

-----Original Message-----
From: Michel Blangy [mailto:mblangy@satco-inc.com]
Sent: Tuesday, January 20, 2009 5:42 PM
To: Seaint@Seaint. Org
Subject: Availability of Older AutoCAD Versions

Is there a legal way to get older versions of AutoCAD like 2004 or 2000? I
found a place that looks slightly suspicious:

http://giojewelry.com/buy_cheap_Autocad_2004.php

Thanks,

Michel

Grouting for Instaltion of Bearings

Dear ALL,

Befor Instaltion of Speherical Bearings, we have to preapre the Goruting pad on the Padestal around 30 mm thick. befor that we have to make trials and set procedure for that.

If anybody have that experiance so please share with me what Procedure shall be followed.

Regards,

Nitin Nagar

Tuesday, January 20, 2009

Re: Lateral Stability of a Box Beam ?

Conrad,
I think that we need to return to the original description of 2 independent
beams with the possibility of adding plywood top and bottom to increase
efficiency of the structure. I did not respond to this since Bill appeared
to have the required approach well considered but was only looking for the
analytical tools to qualify the sheathing.

You raise the special case of single/point symmetric members which require
additional care. I have stated in the past that, in my opinion, purlin
stabilization bracing is an overlooked aspect of building design. I do not
get the impression that this applies to the original post.

I am prepared to discuss L(x,y,z) bracing of cold formed members at great
length but time prevents a rediscovery of published standards.

Per the description of minimizing stable members, the bridging only reduces
the buckle length if there is some added shear resistance mechanism
(stiffness) between the apparent brace points along the length of the
member. That shear path can take the form of sheathing (e.g. plywood, floor
deck, etc., essentially a diagonal link) or a link to anchor or other means
of adding lateral stiffness (re Winter).

Regards
Paul
--
Paul Ransom, P.Eng.
ph 905 639-9628
fax 905 639-3866
ad026@hwcn.org

> From: "Conrad Harrison" <sch.tectonic@bigpond.com>

> What you say makes sense. But I have to think about it, something seems odd.
> Coldformed C/Z are braced by bridging, normally a channel section fastened
> to the web above and below the neutral axis. The bridging provides both
> lateral and torsional restraint: that is Ly, and Lz are reduced.

> Not really about two tied beams. The bridging reduces the half-wavelength of
> the buckle in the thin plate.

> -----Original Message-----
> From: Paul Ransom [mailto:ad026@hwcn.org]
> Sent: Sunday, 18 January 2009 06:54
> To: seaint@seaint.org
> Subject: Re: Lateral Stability of a Box Beam ?
>
> If the 2 beams are identical with identical loading, there is no benefit to
> simple ties. Unless there is some shear resistance mechanism between the
> brace points, the unbraced length for lateral buckling is not reduced and
> the ladder simply behaves like 2 tied beams (e.g. Iy1 + Iy2) and they just
> displace in unison.
>
> Regards
> Paul
> --
> Paul Ransom, P.Eng.
> ph 905 639-9628
> fax 905 639-3866
> ad026@hwcn.org

Re: Availability of Older AutoCAD Versions

Sounds suspicious... Try a authorized reseller through Autodesk...

http://usa.autodesk.com/adsk/servlet/index?siteID=123112&id=1088201

On Tue, Jan 20, 2009 at 3:41 PM, Michel Blangy <mblangy@satco-inc.com> wrote:

Is there a legal way to get older versions of AutoCAD like 2004 or 2000? I
found a place that looks slightly suspicious:

http://giojewelry.com/buy_cheap_Autocad_2004.php

Thanks,

Michel

******* ****** ******* ******** ******* ******* ******* ***
* Read list FAQ at: http://www.seaint.org/list_FAQ.asp
*
* This email was sent to you via Structural Engineers
* Association of Southern California (SEAOSC) server. To
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*
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*
* Questions to seaint-ad@seaint.org. Remember, any email you
* send to the list is public domain and may be re-posted
* without your permission. Make sure you visit our web
* site at: http://www.seaint.org
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--
David Topete, SE

Availability of Older AutoCAD Versions

Is there a legal way to get older versions of AutoCAD like 2004 or 2000? I
found a place that looks slightly suspicious:

http://giojewelry.com/buy_cheap_Autocad_2004.php

Thanks,

Michel

RE: timber pile brace connection

I've seen steel collars but if it's to secure diagonal forces you need a strut along the bottom (unless you deem the ground to act so) and the top (maybe the building does that part), and something to stop the collar from slipping up/down the pile - maybe a collar with interior studs or "teeth" to bite into the pile, you might need to use lag screws ...

Thor A. Tandy P.Eng, C.Eng, Struct.Eng, MIStructE
Victoria, BC
Canada

-----Original Message-----
From: IRV FRUCHTMAN [mailto:ifaeng@yahoo.com]
Sent: Tuesday, January 20, 2009 1:57 PM
To: seaint@seaint.org
Subject: timber pile brace connection

Dear Fellow Engineers
The contractor on a home renovation project wants to use concrete collars around existing timber piles at grade level to secure the pile's diagonal brace rods. (This is to avoid drilling holes thru the piles.) Has anyone seen such a concrete collar, or can suggest other pile connections for this application?
TIA,
Irv

timber pile brace connection

RE: tensile pull-through

The only place I have found this information is in the NASA Fastener Manual by Richard Barrett.

Regards, Harold Sprague

From: akester@cfl.rr.com
To: seaint@seaint.org
Subject: tensile pull-through
Date: Mon, 19 Jan 2009 17:26:51 -0500

I am not sure if I used the right terminology, but I need to calculate the tensile "pull-thru" value for a ¾" diam. A36 all-thread with a nut in tension, with the nut bearing on the steel plate. How do I calculate the pull-thru value? A AISC reference would be great, in the manual or in a design guide.

Thanks in advance, and can you please CC me directly?

Andrew Kester, PE

Orlando, FL

akester@cfl.rr.com

Windows Live™ Hotmail®: Chat. Store. Share. Do more with mail. See how it works.

Windows Live™ Hotmail®: Chat. Store. Share. Do more with mail. Check it out.

Monday, January 19, 2009

Concrete diaphragms

If a concrete diaphragm doesn’t meet the requirements of ASCE7-05 section 12.3 to be idealized as rigid, then what happens? Do you envelope rigid & flexible analyses, as has been suggested for wood? It seems like it should always perform rigidly, even in an all-concrete structure.

thanks,

Gordon Goodell

tensile pull-through

I am not sure if I used the right terminology, but I need to calculate the tensile “pull-thru” value for a ¾” diam. A36 all-thread with a nut in tension, with the nut bearing on the steel plate. How do I calculate the pull-thru value? A AISC reference would be great, in the manual or in a design guide.

Thanks in advance, and can you please CC me directly?

Andrew Kester, PE

Orlando, FL

akester@cfl.rr.com

RE: IBC 2006

Thanks,

That’s what I thought just confirming since the Code was not clear one way or another.

D. Matthew Stuart, P.E., S.E., F.ASCE, SECB

Senior Project Manager

Structural Department

Associate

Engineers and Consultants - CMX

200 Route 9

Manalapan, NJ 07726

732-577-9000 (Ext. 1285)

908-309-8657 (Cell)

732-298-9441 (Fax)

mstuart@CMXEngineering.com

From: Gordon Goodell [mailto:GordonGoodell@harmonydesigninc.com]
Sent: Monday, January 19, 2009 12:24 PM
To: seaint@seaint.org
Subject: RE: IBC 2006

1 side.

regards,

Gordon Goodell

From: Stuart, Matthew [mailto:mStuart@cmxengineering.com]
Sent: Monday, January 19, 2009 10:16 AM
To: seaint@seaint.org
Subject: IBC 2006

Anybody able to confirm if Table 2306.4.5 is based on gyp board on both sides or one side of shear stud wall

D. Matthew Stuart, P.E., S.E., F.ASCE, SECB

RE: IBC 2006

1 side.

regards,

Gordon Goodell

From: Stuart, Matthew [mailto:mStuart@cmxengineering.com]
Sent: Monday, January 19, 2009 10:16 AM
To: seaint@seaint.org
Subject: IBC 2006

Anybody able to confirm if Table 2306.4.5 is based on gyp board on both sides or one side of shear stud wall

D. Matthew Stuart, P.E., S.E., F.ASCE, SECB

IBC 2006

Anybody able to confirm if Table 2306.4.5 is based on gyp board on both sides or one side of shear stud wall

D. Matthew Stuart, P.E., S.E., F.ASCE, SECB

Senior Project Manager

Structural Department

Associate

Engineers and Consultants - CMX

200 Route 9

Manalapan, NJ 07726

732-577-9000 (Ext. 1285)

908-309-8657 (Cell)

732-298-9441 (Fax)

mstuart@CMXEngineering.com

Re: Underground Water Tank in Gulf

Mondo,

In the Gulf region one of your biggest problems, if you have a steel or concrete tank, is corrosion. The soils typically have very high concentrations of evaporated sea salts. Even above ground structural steel is often galvanized then epoxy painted, above ground concrete painted with siloxanes, and below grade concrete foundations are often completely "tanked" with a bithuthene.

Thomas Hunt, S.E.
Fluor

edmondo san jose <engr_mondo@yahoo.com>
01/15/2009 08:49 PM
Please respond to seaint

To	seaint@seaint.org
cc
Subject	Underground Water Tank in Gulf

List,

Any watch-its for the design of large underground potable water tank in gulf region?

Tank size:
length= 50m
width= 100m
height= 5m

TIA,
Mondo

------------------------------------------------------------ The information transmitted is intended only for the person  or entity to which it is addressed and may contain  proprietary, business-confidential and/or privileged material.   If you are not the intended recipient of this message you are  hereby notified that any use, review, retransmission, dissemination,  distribution, reproduction or any action taken in reliance upon  this message is prohibited. If you received this in error, please  contact the sender and delete the material from any computer.    Any views expressed in this message are those of the individual  sender and may not necessarily reflect the views of the company.   ------------------------------------------------------------

Sunday, January 18, 2009

RE: Bridge Seismic Analysis

Jorge,

In general I agree with your objections. Nevertheless the described
procedure is quite common in practice since not every office has expensive
non-linear software available. I believe that the superposition of the two
analysis yields a conservative result. Maybe others more familiar with
bridge design in this forum could clarify better this issue.

Javier Encinas, PE
-----Original Message-----
From: Jorge Jimenez [mailto:joraljim@prtc.net]
Sent: Thursday, January 15, 2009 5:38 PM
To: seaint@seaint.org
Subject: Bridge Seismic Analysis

I'm currently working on a bridge design in Puerto Rico, and I am wondering
about the validity of the procedure for seismic design used for some local
engineers.

As an example, consider a two-span bridge with a central pier. The deck
composed of AASHTO-Girders. At the ends there are counterfort abutments.
There are reinforced elastomeric bearing pads at each support. The bridge
superstructure is intended to be designed as continuous beam, and thus
having positive moments at central pier. There are 2" gaps between the
superstructure and the abutments. There are reinforced concrete shear blocks
in the transverse direction, having 1/2" gaps with the girders.

Structural computer models are:
MODEL 1: A continuous beam and the central pier with springs at
superstructure supports representing the neoprene pads.
MODEL 2: A continuous beam and the central pier with springs at
superstructure supports with 1-spring representing one abutment moving
toward backfill and the remaining springs representing elastomeric pads.

The procedure for seismic modeling and design consider the following steps:
1. Perform the analysis for the gravitary loads.
2. Design the elastomeric pads.
3. Find the equivalent springs to represent the horizontal and vertical
stiffness of the elastomers. The bridge is assumed restrained both
longitudinal and transversely by the elastomers only. No anchor bolts or
another anchoring device is assumed to be used at supports.
4. Perform a multimodal elastic dynamic analysis (MMEDA) for 'Model 1',
considering the elastomers stiffness, both in the horizontal and vertical
direction.
5. Since the horizontal displacements in 'Model 1' are greater than the
bridge gaps, usually in a proportion of 3 to 5, a second MMEDA is made for
'Model 2'. In this new model, the gap at one of the abutments is assumed
closed, and one of the end springs representing the stiffness of the
abutment with the soil passive resistance contribution.
6. Design of the bridge components using the more critical forces from both
analyses.

My objections to the described procedure are:
1. The elastomers were designed only for gravitary loads, but used for
seismic loads and not meeting the specifications for seismic isolated
bridges.
2. The MMEDA assumes systems having free oscillation. Closing gaps produces
collisions disturbing that oscillation and therefore the MMEDA becomes
invalid.
3. The MMEDA assumes the combination of the different vibration modes of the
system. The maximum modal displacements are not simultaneous and there are
some combination procedures (CQC, SRSS and others), to find the total
expected displacements with the contribution of the most significant modes.
Some time interval of free oscillation and certain number of cycles are
required by the system to reach the maximum displacement having a close
correlation with the analytical modal combination techniques. With the
procedure above, it is possible to have one of the end gaps closed even
without the first cycle completion for some modes.
4. Usually the abutments have skew angles respect the bridges longitudinal
axis. Therefore, the dynamic equations of movement are coupled. The gaps
closing and opening in longitudinal and transverse direction behaves
randomly, introducing more non-linearity to the system.
5. A MMEDA is only intended to be used in linear elastic systems. A
horizontal force vs. displacement plot is bilinear. The unloading trajectory
is also bilinear. Is by definition a non-linear system. For non-linear
systems step-by-step analysis or other analyses methods could be used.

My questions:
1. Do you agree with my objections for the described procedure?
2. It is possible to analyze a bridge in the example without using a
non-linear analysis tool?
3. Any comments?

Regards,
Jorge Jimenez, PE