Structural Engineering: 2009-04-19

Friday, April 24, 2009

RE: ACI 318 Appendix D

Another anchor worth considering is the USP DUC anchor. The DUC has been tested per ACI 355.2 for use in seismic and cracked concrete applications, and meets the 2003 IBC and ACI 318-02 Appendix D. They also have good technical support and the ICC ESR's that are necessary for proper installation and special inspection.
http://uspconnectors.com/e_duc.shtml

Regards, Harold Sprague

From: jrgrill@cableone.net
To: seaint@seaint.org
Subject: RE: ACI 318 Appendix D
Date: Thu, 23 Apr 2009 14:59:54 -0700

Bill,

Simpson does have some post installed anchors that will comply with 318 app. D depending on edged distances etc. They also have an undercut anchor called the "Torq-Cut" (The ICC-ES is pending) that will give some very high tension values (22.6 kips for a 5/8" anchor) and give the ductility, however, again there are some edge distance and embedment requirements.

Joe Grill

From: Bill Allen [mailto:t.w.allen@cox.net]
Sent: Tuesday, April 21, 2009 10:22 AM
To: seaint@seaint.org
Subject: RE: ACI 318 Appendix D

...and what ACI 318 Appendix D is saying is that you want to avoid a post installed anchor at all costs!

=8^O

T. William (Bill) Allen, S.E.

ALLEN DESIGNS

Consulting Structural Engineers
V (949) 248-8588 • F(949) 209-2509

-----Original Message-----
From: Andre Sidler [mailto:asidler@hotmail.com]
Sent: Tuesday, April 21, 2009 9:40 AM
To: SEI Listserve
Subject: ACI 318 Appendix D

The reason AISC 341-05 negates the seismic requirements of Appendix D is because the Appenix D requirements are based on the load combinations of ASCE 7-05 (same as ACI 318-05 load combos) which is different than the AISC 341-05 requirements which are based on member capacity demands. Essentially what AISC is saying is that they want the members of the lateral force resisting system to yield before the anchorage to the foundation does. This load is typically much higher than the load combination demands because the member capacity, for instance, is based on RyFyAg. Ry is a factor in AISC 341-05 that increase the yield strength of the memeber based on what the upper bound of the yield strength of the material is, like ASTM A36, etc.

In a nutshell essentially AISC is saying the anchorage must be designed for higher forces than what the load combinations that Appendix D is based on.

I hope that explains it.

I failed to mention the great exception to the AISC requirement and that is the all inclusive load generated by "what the system can deliver" exception.

Andre Sidler, S.E., P.E.
Quantum Consulting Engineers
Seattle, WA

--------------------------------------------------------------------------
9 Message:0009 9
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From: "Joseph R. Grill" <jrgrill@cableone.net>
To: <seaint@seaint.org>
Subject: RE: ACI 318 Appendix D

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I was thinking the same thing. I don't see how AISC can negate ( as William
says ) an ACI code requirement for anchorage in concrete. If there was ever
a problem I don't think I would call ACI as a witness for my defense if I
used the AISC code to do (or not do) the anchorage design, but I bet the
other side would call them in a minute.

Joe Grill

From: William.Sherman@CH2M.com [mailto:William.Sherman@CH2M.com]
Sent: Sunday, April 19, 2009 8:45 AM
To: seaint@seaint.org
Subject: RE: ACI 318 Appendix D

Section 8.5 of AISC 341-05 essentially negates the seismic provisions in ACI
318 Appendix D for anchor bolts at steel column base plates. It is
interesting that the "steel code" allows exceptions to the seismic
provisions of the "concrete code"!?

Bill Sherman

CH2M HILL / DEN

720-286-2792

_____

From: Jules [mailto:JulesG@socal.rr.com]
Sent: Sunday, April 19, 2009 9:38 AM
To: seaint@seaint.org
Subject: Re: ACI 318 Appendix D

AISC 341-05 is better known as "Seismic Provisions for Structural Steel
Buildings"

----- Original Message -----

From: Larry Hauer <mailto:lhauer@live.com>

To: Struct. Eng. Assoc. <mailto:seaint@seaint.org>

Sent: Sunday, April 19, 2009 8:16 AM

Subject: RE: ACI 318 Appendix D

Jules,

What is ACI 341-05? It appears to be a manual I don't have. Can you tell us
what Section 8.5 says?

Thanks in Advance

Larry Hauer S.E.

_____

From: JulesG@socal.rr.com
To: seaint@seaint.org
Subject: Re: ACI 318 Appendix D
Date: Sat, 18 Apr 2009 08:39:21 -0700

Bill,

I've been reading through all the answers on this subject but nobody
mentioned the exception on AISC 341-05 8.5. It solves most of the problems
for anchoring column base plates, I use it.

Regards.

Jules

----- Original Message -----

From: Bill Allen

To: Seaint

Sent: Thursday, April 16, 2009 1:25 PM

Subject: ACI 318 Appendix D

O.K., I've beaten my head against the wall long enough. I've decided that a
post-installed anchor won't work in tension for anything I design.

I feel better now.

T. William (Bill) Allen, S.E.

ALLEN DESIGNS

Consulting Structural Engineers
V (949) 248-8588 . F(949) 209-2509

Rediscover Hotmail®: Get e-mail storage that grows with you. Check it out.

RE: RISA-3D Steel WF Torsion

Bill,
Wide flange shapes have a very low capacity in torsion. The other issue with torsion is the end connections. If the local member is not torsionally restrained at the support, there is no resistance, but that normally will give you an instability. I would still be suspicious of a problem with the member end contraints / releases.

Regards, Harold Sprague

> From: bill@polhemus.cc [mailto:bill@polhemus.cc]
> Sent: Thursday, April 16, 2009 1:57 PM
> To: seaint@seaint.org
> Subject: RISA-3D Steel WF Torsion
>
> I'm getting some anomalous results from a rather complicated steel frame run
> on RISA-3D. I'm doing RSA with this, and I have already noticed some
> peculiar results (at least for this "statics-based" puke).
>
> When I look at the "Shear Check" plot on-screen, I'll see a lot of failures
> of wide-flange members (mostly beams), many with rather large "understrength
> factors" (if you will; I guess we used to say "overstress"), between 3.0 and
> 4.0.
>
> But when I look at the specifics of a given member, I don't see any large
> shears at all. I'm inclined to think it might be a torsion problem showing
> up, but even the torsion doesn't seem that great (although it is present).
>
> Any thoughts?
>

Rediscover Hotmail®: Get e-mail storage that grows with you. Check it out.

Thursday, April 23, 2009

RE: RISA-3D Steel WF Torsion

Bill -

You will want to review the Member Torsion spreadsheet results to get a
better indication of what shear stresses are being induced in the member
from the torsional loading. Whenever wide flange members are failing in
shear, this is the first place that I would look.

Sincerely,

Josh Plummer, SE

RISA Technologies

-----------------
From: "Alex Nacionales" <alex1961@telus.net>
To: <seaint@seaint.org>
Subject: RE: RISA-3D Steel WF Torsion

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Try checking your frame model and make sure that all columns supporting the
beams which are failing have the correct support condition (fixed or
pinned). I had a similar problem when I forgot to put a support condition to
column below a beam.

_____

From: bill@polhemus.cc [mailto:bill@polhemus.cc]
Sent: Thursday, April 16, 2009 1:57 PM
To: seaint@seaint.org
Subject: RISA-3D Steel WF Torsion

I'm getting some anomalous results from a rather complicated steel frame run
on RISA-3D. I'm doing RSA with this, and I have already noticed some
peculiar results (at least for this "statics-based" puke).

When I look at the "Shear Check" plot on-screen, I'll see a lot of failures
of wide-flange members (mostly beams), many with rather large "understrength
factors" (if you will; I guess we used to say "overstress"), between 3.0 and
4.0.

But when I look at the specifics of a given member, I don't see any large
shears at all. I'm inclined to think it might be a torsion problem showing
up, but even the torsion doesn't seem that great (although it is present).

Any thoughts?

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RE: ACI 318 Appendix D

Bill,

Simpson does have some post installed anchors that will comply with 318 app. D depending on edged distances etc. They also have an undercut anchor called the “Torq-Cut” (The ICC-ES is pending) that will give some very high tension values (22.6 kips for a 5/8” anchor) and give the ductility, however, again there are some edge distance and embedment requirements.

Joe Grill

From: Bill Allen [mailto:t.w.allen@cox.net]
Sent: Tuesday, April 21, 2009 10:22 AM
To: seaint@seaint.org
Subject: RE: ACI 318 Appendix D

...and what ACI 318 Appendix D is saying is that you want to avoid a post installed anchor at all costs!

=8^O

T. William (Bill) Allen, S.E.

ALLEN DESIGNS

Consulting Structural Engineers
V (949) 248-8588 • F(949) 209-2509

-----Original Message-----
From: Andre Sidler [mailto:asidler@hotmail.com]
Sent: Tuesday, April 21, 2009 9:40 AM
To: SEI Listserve
Subject: ACI 318 Appendix D

I Need California SE References

List,

I am, and have been for the last 4+ years, a licensed civil/ structural engineer in Ohio (which has a generic PE certification). I have passed the NCEES SEI and NCEES SEII exams and have a Civil PE in California and Ohio. I have been involved in structural design mostly in industrial facilities across the US (under the supervision of my boss, an SE in California), but I have a broad range of experience in design in high wind and moderate seismic areas.

My predicament is that I live in Ohio and the only California SE that I know in Ohio who is familiar with my work is my boss and that many of the projects that I am involved in would classify as "other structures" or equipment platforms. California requires references from 3 SE's licensed in California. I understand that I could gain references by having someone look over samples of my work and/or talking with me and then judging that he/she feels I am qualified to sit for this test. Would anyone with a California SE be willing to do this?

I put together a few sets of previously submitted calculations demonstrating experience in the areas required by the state of California and am more than willing to send them out to anyone willing to look over them in order that I can use them for a reference. I am registering for the October 2009 exam and the deadline for registration in July 17th. If someone will look over my calculations and respond back to me before July, I can probably still make the deadline and take the test.

Please contact me privately if you're willing. Thanks.

Brian S Bossley, P.E.
7610 Olentangy River Rd
Columbus, OH 43235
(614) 847-1110 x 121 (ph)
(614) 847-1116 (fax)

RE: RISA-3D Steel WF Torsion

Bill,

If you could send me your model I would be happy to look at it.

Thanks,

Bruce Bates
RISA Technologies

_____

From: bill@polhemus.cc [mailto:bill@polhemus.cc]
Sent: Thursday, April 16, 2009 1:57 PM
To: seaint@seaint.org
Subject: RISA-3D Steel WF Torsion

Any thoughts?

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Try checking your frame model and =
make
sure that all columns supporting the beams which are failing have the =
correct
support condition (fixed or pinned). I had a similar problem when I =
forgot to
put a support condition to column below a =
beam.<o:p></o:p>

<o:p> </o:p>

<div>

</div>

From: =
bill@polhemus.cc
[mailto:bill@polhemus.cc] 
Sent: Thursday, April 16, =
2009
1:57 PM 
To: <st1:PersonName =
w:st=3D"on">seaint@seaint.org</st1:PersonName> 
Subject: RISA-3D Steel WF =
Torsion<o:p></o:p>

</div>

<o:p> </o:p>

I'm
getting some anomalous results from a rather complicated steel frame run =
on
RISA-3D. I'm doing RSA with this, and I have already noticed some =
peculiar
results (at least for this "statics-based" =
puke).<o:p></o:p>

When I
look at the "Shear Check" plot on-screen, I'll see a lot of =
failures
of wide-flange members (mostly beams), many with rather large
"understrength factors" (if you will; I guess we used to say
"overstress"), between 3.0 and =
4.0.<o:p></o:p>

But when
I look at the specifics of a given member, I don't see any large shears =
at all.
I'm inclined to think it might be a torsion problem showing up, but even =
the
torsion doesn't seem that great (although it is =
present).<o:p></o:p>

Any thoughts?<o:p></o:p>

</div>

</body>

</html>

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Wednesday, April 22, 2009

RE: RISA-3D Steel WF Torsion

Try checking your frame model and make sure that all columns supporting the beams which are failing have the correct support condition (fixed or pinned). I had a similar problem when I forgot to put a support condition to column below a beam.

From: bill@polhemus.cc [mailto:bill@polhemus.cc]
Sent: Thursday, April 16, 2009 1:57 PM
To: seaint@seaint.org
Subject: RISA-3D Steel WF Torsion

I'm getting some anomalous results from a rather complicated steel frame run on RISA-3D. I'm doing RSA with this, and I have already noticed some peculiar results (at least for this "statics-based" puke).

When I look at the "Shear Check" plot on-screen, I'll see a lot of failures of wide-flange members (mostly beams), many with rather large "understrength factors" (if you will; I guess we used to say "overstress"), between 3.0 and 4.0.

But when I look at the specifics of a given member, I don't see any large shears at all. I'm inclined to think it might be a torsion problem showing up, but even the torsion doesn't seem that great (although it is present).

Any thoughts?

Earthquake

More about earthquakes:

Earthquake simulation for So Cal under a 7.8M San Andreas rupture (video is in real time but without audio):

http://www.youtube.com/watch?v=xioHswbahPc

This video is twice as fast as real time with a different perspective, but has audio: http://www.scivee.tv/node/3076

V. Steve Gordin, SE
Irvine CA

L'Aquila Earthquake info

http://www.eqclearinghouse.org/italy-090406/

RE: Steel strsss increase for seismic loads

Wood technically is not really the same as the 1/3 stress increase...it is a load duration factor...even if it produces a similar "effect".

I will further note that some local jurisdictions further restrict the load duration factors.

Regards,

Scott

Adrian, MI

From: Larry Hauer [mailto:lhauer@live.com]
Sent: Tuesday, April 21, 2009 5:36 PM
To: Struct. Eng. Assoc.
Subject: RE: Steel strsss increase for seismic loads

Thanks Tom, that is what I thought with WOOD and masonry the only materials being allowed the 1/3 increase. (Oh well!).

Larry

To: seaint@seaint.org
Subject: Re: Steel strsss increase for seismic loads
From: Tom.Hunt@fluor.com
Date: Tue, 21 Apr 2009 14:22:48 -0700

Larry,

If you use the load combinations from IBC Section 1605.3.1 then a 1/3 increase in allowable stresses is not permitted. If you use the IBC load combinations from Section 1605.3.2 then you are allowed to use the 1/3 increases but ONLY if allowed by the material chapter. Unfortunately for steel neither the IBC nor AISC have the 1/3 increase factor anymore. I believe the only material left is masonry and in some cases geotechnical soil bearing.

Thomas Hunt, S.E.
Fluor

Larry Hauer <lhauer@live.com>
04/21/2009 02:09 PM
Please respond to seaint

To	"Struct. Eng. Assoc." <seaint@seaint.org>
cc
Subject	Steel strsss increase for seismic loads

I am checking existing steel roof framing members for support of replaced roof top mechanical equipment. Obviously, I need to design per the '06 IBC/'07 CBC for seismic forces and will be using ASD equation 16-20 of the IBC/CBC. Based on Sec. 1605.3.1.1 it looks like a 1/3 increase in allowable stresses for the existing steel roof members is no longer allowed when considering seismic loads. Is this correct?

Thanks in advance,

Larry Hauer

Windows Live™ Hotmail®:…more than just e-mail. Check it out.

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Re: Light gage boxed header limitations

You're more likely to get shear buckling / web crippling than LT buckling. LT is an issue when sections have a low torsional rigidity, but for a properly designed/built box beam you should have a surplus of torsional moment capacity unless you have a very deep header.

BTW do you have to design something or can you just call a vendor like Metwood (who makes CFS box headers with rebar added for additional strength) and give them the loading?

Jordan

Jim Wilson wrote:

At some span, I would think that 6" wide light gage box beams would become susceptible to lateral-torsional buckling where there are no lateral brace points along its length. Light gage framing bending and compression members are typically locked into sheathed wall/floor/roof systems and I just have to question if there is a way to address this in accordance with the AISI Specs for header design.

Jim

From: Daryl Richardson <h.d.richardson@shaw.ca>
To: seaint@seaint.org
Sent: Tuesday, April 21, 2009 12:49:22 PM
Subject: Re: Light gage boxed header limitations

Jim,

 Vertical load shouldn't be a problem; you can always make a light truss.

 For horizontal you might check deflection; see how it compares with a limit of say half an inch. That is where I think you might have problems.

Regards,

H. Daryl Richardson

----- Original Message -----

From: Jim Wilson

To: seaint@seaint.org

Sent: Tuesday, April 21, 2009 6:45 AM

Subject: Light gage boxed header limitations

I have been asked by an EOR to design a 24ft long boxed header with light gage framing. This member will be laterally unbraced in a curtain wall assembly. Are there any practical limits to the span of such a header in a 6" wide wall? The loads are only in the 200lb/ft range vertical. Horizontal loads are less.

The "Standard for Cold-Formed Steel Framing - Header Design" by AISI only states length limits for L-shaped headers. Header calcs in this standard only address bending, shear and web crippling. I don't believe that the AISI main spec will adequately model a long-span composite assembly.

I would like to either find a conclusive way to design this header or prove that it is not practical.

Thanks in advance.

Jim Wilson, PE

Stroudsburg, PA

RE: Steel strsss increase for seismic loads

It's "built in" now.

Windows Live™ Hotmail®:…more than just e-mail. Check it out.

Re: Dubai Wind Speed requirements

Return Receipt

Your Re: Dubai Wind Speed requirements
document:

was Tom.Hunt@fluor.com
received
by:

at: 04/22/2009 07:10:00 PDT

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The information transmitted is intended only for the person
or entity to which it is addressed and may contain
proprietary, business-confidential and/or privileged material.
If you are not the intended recipient of this message you are
hereby notified that any use, review, retransmission, dissemination,
distribution, reproduction or any action taken in reliance upon
this message is prohibited. If you received this in error, please
contact the sender and delete the material from any computer.

Any views expressed in this message are those of the individual
sender and may not necessarily reflect the views of the company.
------------------------------------------------------------

Re: Light gage boxed header limitations

At some span, I would think that 6" wide light gage box beams would become susceptible to lateral-torsional buckling where there are no lateral brace points along its length. Light gage framing bending and compression members are typically locked into sheathed wall/floor/roof systems and I just have to question if there is a way to address this in accordance with the AISI Specs for header design.

Jim

From: Daryl Richardson <h.d.richardson@shaw.ca>
To: seaint@seaint.org
Sent: Tuesday, April 21, 2009 12:49:22 PM
Subject: Re: Light gage boxed header limitations

Jim,

Vertical load shouldn't be a problem; you can always make a light truss.

For horizontal you might check deflection; see how it compares with a limit of say half an inch. That is where I think you might have problems.

Regards,

H. Daryl Richardson

----- Original Message -----

From: Jim Wilson

To: seaint@seaint.org

Sent: Tuesday, April 21, 2009 6:45 AM

Subject: Light gage boxed header limitations

I have been asked by an EOR to design a 24ft long boxed header with light gage framing. This member will be laterally unbraced in a curtain wall assembly. Are there any practical limits to the span of such a header in a 6" wide wall? The loads are only in the 200lb/ft range vertical. Horizontal loads are less.

The "Standard for Cold-Formed Steel Framing - Header Design" by AISI only states length limits for L-shaped headers. Header calcs in this standard only address bending, shear and web crippling. I don't believe that the AISI main spec will adequately model a long-span composite assembly.

I would like to either find a conclusive way to design this header or prove that it is not practical.

Thanks in advance.

Jim Wilson, PE

Stroudsburg, PA

Re: Dubai Wind Speed requirements

Hello Paul,

we've designed a large multypurpose stadium in Dubai. The specifications for

wind loads were given by the British Standard BS 6399, part 2 with 45 m/s.

In Dubai in most cases the BS is in use.

Rico Harzer S.E.

Germany

Tuesday, April 21, 2009

Re: Dubai Wind Speed requirements

dubai wind speed

Basic wind speed (hourly mean value) Vb is 26m/s

which is equivalent to 3s gust value of 45m/s as per cp111

Mariyamma

----- Original Message -----

From: Paul Feather

To: seaint@seaint.org

Sent: Tuesday, April 21, 2009 12:33 AM

Subject: Dubai Wind Speed requirements

I am looking for the wind speed requirements in Dubai. Can anyone provide some assistance?

Paul Feather PE, SE

pfeather@SE-Solutions.net

www.SE-Solutions.net

RE: Steel strsss increase for seismic loads

According to Breyers section 2.17, stress increase is not allowed for wood.
Duration of load Cd for ASD and time effect factor lambda for LFRD are
applied as modifications, but might be allowable per IBC ASD alternate load
combinations.

-----Original Message-----
From: Drew Morris [mailto:dmorris@bbfm.com]
Sent: Tuesday, April 21, 2009 3:05 PM
To: seaint@seaint.org
Subject: Re: Steel strsss increase for seismic loads

I believe wood uses the duration factor Cd so it can vary from 0.9 to 1.6.

Larry Hauer wrote:
> Thanks Tom, that is what I thought with WOOD and masonry the only
> materials being allowed the 1/3 increase. (Oh well!).
>
> Larry
>
> ----------------------------------------------------------------------
> --
> To: seaint@seaint.org
> Subject: Re: Steel strsss increase for seismic loads
> From: Tom.Hunt@fluor.com
> Date: Tue, 21 Apr 2009 14:22:48 -0700
>
>
> Larry,
>
> If you use the load combinations from IBC Section 1605.3.1 then a 1/3
> increase in allowable stresses is not permitted. If you use the IBC
> load combinations from Section 1605.3.2 then you are allowed to use
> the 1/3 increases *but ONLY if allowed by the material chapter*.
> Unfortunately for steel neither the IBC nor AISC have the 1/3
> increase factor anymore. I believe the only material left is masonry
> and in some cases geotechnical soil bearing.
>
> Thomas Hunt, S.E.
> Fluor
>
>
>
>
>
> *Larry Hauer <lhauer@live.com>*
> 04/21/2009 02:09 PM
> Please respond to seaint
> To
> "Struct. Eng. Assoc." <seaint@seaint.org> cc
>
> Subject
> Steel strsss increase for seismic loads
>
>
>
>
>
>
>
>
>
> I am checking existing steel roof framing members for support of
> replaced roof top mechanical equipment. Obviously, I need to design
> per the '06 IBC/'07 CBC for seismic forces and will be using ASD
> equation 16-20 of the IBC/CBC. Based on Sec. 1605.3.1.1 it looks like
> a 1/3 increase in allowable stresses for the existing steel roof
> members is no longer allowed when considering seismic loads. Is this
> correct?
>
> Thanks in advance,
>
> Larry Hauer
>
> ----------------------------------------------------------------------
> -- Windows LiveT HotmailR:.more than just e-mail. _Check it out._
> <http://windowslive.com/online/hotmail?ocid=TXT_TAGLM_WL_HM_more_04200
> 9>
> ------------------------------------------------------------
> The information transmitted is intended only for the person or entity
> to which it is addressed and may contain proprietary,
> business-confidential and/or privileged material.
> If you are not the intended recipient of this message you are hereby
> notified that any use, review, retransmission, dissemination,
> distribution, reproduction or any action taken in reliance upon this
> message is prohibited. If you received this in error, please contact
> the sender and delete the material from any computer.
>
> Any views expressed in this message are those of the individual sender
> and may not necessarily reflect the views of the company.
> ------------------------------------------------------------
>
>
> ------------------------------------------------------------------------
> Rediscover HotmailR: Get quick friend updates right in your inbox.
> Check it out.
>
<http://windowslive.com/RediscoverHotmail?ocid=TXT_TAGLM_WL_HM_Rediscover_Up
dates2_042009>

Re: Steel strsss increase for seismic loads

I believe wood uses the duration factor Cd so it can vary from 0.9 to 1.6.

Larry Hauer wrote:
> Thanks Tom, that is what I thought with WOOD and masonry the only
> materials being allowed the 1/3 increase. (Oh well!).
>
> Larry
>
> ------------------------------------------------------------------------
> To: seaint@seaint.org
> Subject: Re: Steel strsss increase for seismic loads
> From: Tom.Hunt@fluor.com
> Date: Tue, 21 Apr 2009 14:22:48 -0700
>
>
> Larry,
>
> If you use the load combinations from IBC Section 1605.3.1 then a 1/3
> increase in allowable stresses is not permitted. If you use the IBC
> load combinations from Section 1605.3.2 then you are allowed to use
> the 1/3 increases *but ONLY if allowed by the material chapter*.
> Unfortunately for steel neither the IBC nor AISC have the 1/3
> increase factor anymore. I believe the only material left is masonry
> and in some cases geotechnical soil bearing.
>
> Thomas Hunt, S.E.
> Fluor
>
>
>
>
>
> *Larry Hauer <lhauer@live.com>*
> 04/21/2009 02:09 PM
> Please respond to seaint
> To
> "Struct. Eng. Assoc." <seaint@seaint.org>
> cc
>
> Subject
> Steel strsss increase for seismic loads
>
>
>
>
>
>
>
>
>
> I am checking existing steel roof framing members for support of
> replaced roof top mechanical equipment. Obviously, I need to design
> per the '06 IBC/'07 CBC for seismic forces and will be using ASD
> equation 16-20 of the IBC/CBC. Based on Sec. 1605.3.1.1 it looks like
> a 1/3 increase in allowable stresses for the existing steel roof
> members is no longer allowed when considering seismic loads. Is this
> correct?
>
> Thanks in advance,
>
> Larry Hauer
>
> ------------------------------------------------------------------------
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RE: Steel strsss increase for seismic loads

Thanks Tom, that is what I thought with WOOD and masonry the only materials being allowed the 1/3 increase. (Oh well!).

Larry

Larry Hauer <lhauer@live.com>
04/21/2009 02:09 PM
Please respond to seaint

To	"Struct. Eng. Assoc." <seaint@seaint.org>
cc
Subject	Steel strsss increase for seismic loads

Windows Live™ Hotmail®:…more than just e-mail. Check it out.

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RE: Steel strsss increase for seismic loads

Don't forget wood.

Dave Puskas, PE, SECB

(Senior Engineer, Structural)

Cyntergy aec

320 South Boston 12th Floor

Tulsa, Oklahoma 74103

918-877-6000 ext. 394

From: Tom.Hunt@fluor.com [mailto:Tom.Hunt@fluor.com]
Sent: Tuesday, April 21, 2009 4:23 PM
To: seaint@seaint.org
Subject: Re: Steel strsss increase for seismic loads

Larry,

If you use the load combinations from IBC Section 1605.3.1 then a 1/3 increase in allowable stresses is not permitted. If you use the IBC load combinations from Section 1605.3.2 then you are allowed to use the 1/3 increases but ONLY if allowed by the material chapter. Unfortunately for steel neither the IBC nor AISC have the 1/3 increase factor anymore. I believe the only material left is masonry and in some cases geotechnical soil bearing.

Thomas Hunt, S.E.
Fluor

Larry Hauer <lhauer@live.com>
04/21/2009 02:09 PM
Please respond to seaint

To	"Struct. Eng. Assoc." <seaint@seaint.org>
cc
Subject	Steel strsss increase for seismic loads

Windows Live™ Hotmail®:…more than just e-mail. Check it out.

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Re: Steel strsss increase for seismic loads

Larry Hauer <lhauer@live.com>
04/21/2009 02:09 PM
Please respond to seaint

To	"Struct. Eng. Assoc." <seaint@seaint.org>
cc
Subject	Steel strsss increase for seismic loads

Windows Live™ Hotmail®:…more than just e-mail. Check it out.

------------------------------------------------------------ The information transmitted is intended only for the person  or entity to which it is addressed and may contain  proprietary, business-confidential and/or privileged material.   If you are not the intended recipient of this message you are  hereby notified that any use, review, retransmission, dissemination,  distribution, reproduction or any action taken in reliance upon  this message is prohibited. If you received this in error, please  contact the sender and delete the material from any computer.    Any views expressed in this message are those of the individual  sender and may not necessarily reflect the views of the company.   ------------------------------------------------------------

Steel strsss increase for seismic loads

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Re: ACI 318 Appendix D

No I haven't heard of any testing, but it would be welcome provided they get common geometries for various regions. We say a small curse when someone asks to put a 12" x10' shear wall into a residence, and we find out they've already poured a 6" stemwall, or better yet have a CMU basement wall they'd like to anchor to.

Jordan

Jeff Smith wrote:

I have been reviewing the appendix D anchorage issues and in particular as it relates to Simpson Strongwall holdown anchors. Using Appendix D for holdown anchors along a property line stem wall "L" footing results in inadequate capacities. I spoke with a Simpson Strong-Tie engineer about the dilemma. It was my understanding that Simpson is going to do full scale testing of their strong wall assemblies anchored in small concrete edge distance conditions that will result in approved anchorage capacities greater than what would be determined from appendix D. I am not sure if this is true, has anyone else discussed this issue with Simpson?

Jeff

From: Bill Allen [mailto:t.w.allen@cox.net]
Sent: Tuesday, April 21, 2009 10:22 AM
To: seaint@seaint.org
Subject: RE: ACI 318 Appendix D

...and what ACI 318 Appendix D is saying is that you want to avoid a post installed anchor at all costs!

=8^O

T. William (Bill) Allen, S.E.

ALLEN DESIGNS

Consulting Structural Engineers
V (949) 248-8588 • F(949) 209-2509

-----Original Message-----
From: Andre Sidler [mailto:asidler@hotmail.com]
Sent: Tuesday, April 21, 2009 9:40 AM
To: SEI Listserve
Subject: ACI 318 Appendix D

The reason AISC 341-05 negates the seismic requirements of Appendix D is because the Appenix D requirements are based on the load combinations of ASCE 7-05 (same as ACI 318-05 load combos) which is different than the AISC 341-05 requirements which are based on member capacity demands. Essentially what AISC is saying is that they want the members of the lateral force resisting system to yield before the anchorage to the foundation does. This load is typically much higher than the load combination demands because the member capacity, for instance, is based on RyFyAg. Ry is a factor in AISC 341-05 that increase the yield strength of the memeber based on what the upper bound of the yield strength of the material is, like ASTM A36, etc.

In a nutshell essentially AISC is saying the anchorage must be designed for higher forces than what the load combinations that Appendix D is based on.

I hope that explains it.

I failed to mention the great exception to the AISC requirement and that is the all inclusive load generated by "what the system can deliver" exception.

Andre Sidler, S.E., P.E.
Quantum Consulting Engineers
Seattle, WA

--------------------------------------------------------------------------
9 Message:0009 9
--------------------------------------------------------------------------
From: "Joseph R. Grill" <jrgrill@cableone.net>
To: <seaint@seaint.org>
Subject: RE: ACI 318 Appendix D

This is a multipart message in MIME format.

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Content-Type: text/plain;
charset="us-ascii"
Content-Transfer-Encoding: 7bit

I was thinking the same thing. I don't see how AISC can negate ( as William
says ) an ACI code requirement for anchorage in concrete. If there was ever
a problem I don't think I would call ACI as a witness for my defense if I
used the AISC code to do (or not do) the anchorage design, but I bet the
other side would call them in a minute.

Joe Grill

From: William.Sherman@CH2M.com [mailto:William.Sherman@CH2M.com]
Sent: Sunday, April 19, 2009 8:45 AM
To: seaint@seaint.org
Subject: RE: ACI 318 Appendix D

Section 8.5 of AISC 341-05 essentially negates the seismic provisions in ACI
318 Appendix D for anchor bolts at steel column base plates. It is
interesting that the "steel code" allows exceptions to the seismic
provisions of the "concrete code"!?

Bill Sherman

CH2M HILL / DEN

720-286-2792

_____

From: Jules [mailto:JulesG@socal.rr.com]
Sent: Sunday, April 19, 2009 9:38 AM
To: seaint@seaint.org
Subject: Re: ACI 318 Appendix D

AISC 341-05 is better known as "Seismic Provisions for Structural Steel
Buildings"

----- Original Message -----

From: Larry Hauer <mailto:lhauer@live.com>

To: Struct. Eng. Assoc. <mailto:seaint@seaint.org>

Sent: Sunday, April 19, 2009 8:16 AM

Subject: RE: ACI 318 Appendix D

Jules,

What is ACI 341-05? It appears to be a manual I don't have. Can you tell us
what Section 8.5 says?

Thanks in Advance

Larry Hauer S.E.

_____

From: JulesG@socal.rr.com
To: seaint@seaint.org
Subject: Re: ACI 318 Appendix D
Date: Sat, 18 Apr 2009 08:39:21 -0700

Bill,

I've been reading through all the answers on this subject but nobody
mentioned the exception on AISC 341-05 8.5. It solves most of the problems
for anchoring column base plates, I use it.

Regards.

Jules

----- Original Message -----

From: Bill Allen

To: Seaint

Sent: Thursday, April 16, 2009 1:25 PM

Subject: ACI 318 Appendix D

O.K., I've beaten my head against the wall long enough. I've decided that a
post-installed anchor won't work in tension for anything I design.

I feel better now.

T. William (Bill) Allen, S.E.

ALLEN DESIGNS

Consulting Structural Engineers
V (949) 248-8588 . F(949) 209-2509

RE: ACI 318 Appendix D

I have been reviewing the appendix D anchorage issues and in particular as it relates to Simpson Strongwall holdown anchors. Using Appendix D for holdown anchors along a property line stem wall "L" footing results in inadequate capacities. I spoke with a Simpson Strong-Tie engineer about the dilemma. It was my understanding that Simpson is going to do full scale testing of their strong wall assemblies anchored in small concrete edge distance conditions that will result in approved anchorage capacities greater than what would be determined from appendix D. I am not sure if this is true, has anyone else discussed this issue with Simpson?

Jeff

From: Bill Allen [mailto:t.w.allen@cox.net]
Sent: Tuesday, April 21, 2009 10:22 AM
To: seaint@seaint.org
Subject: RE: ACI 318 Appendix D

...and what ACI 318 Appendix D is saying is that you want to avoid a post installed anchor at all costs!

=8^O

T. William (Bill) Allen, S.E.

ALLEN DESIGNS

Consulting Structural Engineers
V (949) 248-8588 • F(949) 209-2509

-----Original Message-----
From: Andre Sidler [mailto:asidler@hotmail.com]
Sent: Tuesday, April 21, 2009 9:40 AM
To: SEI Listserve
Subject: ACI 318 Appendix D