Structural Engineering: 2007-09-09

Saturday, September 15, 2007

Re: Minimum RC column dimension, rebars

Alex,

I don't see any exception to UBC 1921.4.1.1 for seismic induced forces. You really have to use 12" min. dimension. Now if the factored axial load is less than 0.10Ag*f'c then the member becomes a flexural member. In this case you are bound to follow UBC 1921.3.1.4 which permits minimum width of 10".

Column is defined as a member with a ratio of height to least lateral dimension of 3 or greater and primarily to support axial compressive load. So it does not mean that when the member is vertical it is called a column. The load define the member if it as column or a beam. A beam might be horizontal but if it is subjected primarily to axial compression then it becomes a column and vice versa. Therefore a member subjected to less than .10Ag*f'c is considered to be a flexural member or beam.

I don't see any provision that limits stirrups and ties to at least 3/8" dia. The only requirement per UBC 1903.5.1 is that reinforcement shall be deformed bars. I remember from the table of an old book by Harry Parker that #2 rebar comes in plain bars only. This is the reason why most people use #3 (3/8" dia.). So in my opinion if #2 is already available in deformed bars then we can use that too. Anyway, the lesser the area the closer the spacing is.

Hope this will help you.

Alfonso S. Quilala Jr., P.E.

----- Original Message ----
From: Alex C. Nacionales <anacionales@insightts.net>
To: seaint@seaint.org
Sent: Friday, September 14, 2007 8:14:05 AM
Subject: Minimum RC column dimension, rebars

For earthquake resistant structures, RC columns are limited

to a minimum dimension of 12 in. or 305mm. , UBC 1997 1921.4.1.1 . and a minimum steel ratio of .01

For two story residential buildings with column spacing of

4.0 meters or less and height less than 7.5 meters , this results to columns with strengths in excess of what is needed to carry the required loads.

Is there any exception to this code provision that will allow an 8" or 10" RC column.?

If the factored axial load is less than 0.10Ag*Fc' will this permit me to use a 10" RC column.?

Where can I find the the code that limits stirrups and ties

to at least 3/8" ~(10mm) Dia.

Thanks in Advance.

Alex Nacionales, C.E.

Fussy? Opinionated? Impossible to please? Perfect. Join Yahoo!'s user panel and lay it on us.

Friday, September 14, 2007

ASCE Transactions Paper N. 1835

Gilvan,

Try the Earthquake Engineering library that UC Berkeley maintains. It can
be found at http://nisee.berkeley.edu/

Good Luck.

David A. Topete, SE

RE: Retaining Wall Questions

On 13 Sep 2007 at 17:56, Wesley Werner wrote:

> Jordan,

> This doesn't exactly answer your question, but at a seminar I

> attended on segmental retaining walls the

> presenter said that you had to design the walls as one wall unless the

> horizontal distance between them was at least twice the height of the

> lower wall. He didn't give us the theory behind the rule-of-thumb,

> though. Also, he said that you needed to check the global stability of

> the whole tiered system. He said that he had seen at least one case

> where a whole hillside gave way behind the walls rather than the walls

> themselves giving way.

> Wesley C. Werner

Hello,

I've bolded the last two sentences in the above quote, because it touches

on what I was going to say about these designs.

Global stability is part of the reason for the setbacks and not just to

develop the passive pressures. Checking the overall slope stability is

a very important thing. If the slope failure plane in below all of your walls

and daylights on the downhill side of your lowest wall, then you risk

having the whole thing come down walls and all.

Check the slope and make sure that you've extended the wall below

that slip plane, and designed for the loads that the slope may impart

to the wall if/when it tries to slip.

Take Care,

Lloyd

RE: embedded plate

You might try these plastic clips for flush mounts in masonry. I suspect, more often that not, the plate will not align with the coursework and the contractor will end up knocking out the face shell and webs and attaching the weld plate to a plywood form.

http://www.tricomasonryproducts.com/

Christopher Banbury, PE

President

Ark Engineering, Inc.

PO Box 10129, Brooksville, FL 34603

22 North Broad ST, Brooksville, FL 34601

Phone: (352) 754-2424

Fax: (352) 754-2412

www.arkengineering.net

From: Jason Atwood [mailto:atwoodjason@yahoo.com]
Sent: Friday, September 14, 2007 11:01 AM
To: seaint@seaint.org
Subject: embedded plate

Speaking of embedded plates how do contractors set these in their forms-do they glue it with PL or something?

Jason

RE: Practical Example of wind design for multi-story residential wood framed analysis needed for 2006 IBC and ASCE 7-05

Dennis and Bill:

In my opinion the Wind Analysis based on CBC 2007 (ASCE 7-05) for low and mid rise structures are relatively easy but a lot more time consuming than the current CBC 2001.
However based on the information that I've got form Howard Smith & Bill Staelhlin down at the DSA, about two weeks ago, a new simplified Wind Design method will be available by early 2008 for California structural engineers.

Regards

Khashayar (Casey) Hemmatyar, SE
Private e: khemmatyar@hotmail.com

________________________________________
From: Bill Polhemus [mailto:bill@polhemus.cc]
Sent: Thursday, September 13, 2007 7:29 PM
To: seaint@seaint.org
Subject: Re: Practical Example of wind design for multi-story residential wood framed analysis needed for 2006 IBC and ASCE 7-05

Dennis Wish wrote:
For the first time in history, the seismic design is so much easier to understand than the wind loads but I suppose this makes sense in the coastal regions subject to high wind loads.
"You said a mouthful!"

(Signed)
Beaumont, Texas
(H. Rita, 9/2005; H. Humberto, 9/2007)

(N.B. Dennis, go here. The wind load provisions of 7-05 are not very different at all from those of 7-02 - I know, I asked Dr. Mehta if he was going to write a guide for 7-05 and he told me "there's no need, not enough changed.")

________________________________________
From: Dennis Wish [mailto: dennis.wish@verizon.net]
Sent: Thursday, September 13, 2007 5:42 PM
To: seaint@seaint.org
Subject: Practical Example of wind design for multi-story residential wood framed analysis needed for 2006 IBC and ASCE 7-05

I have a lot of information related to the calculations ....

RE: Practical Example of wind design for multi-story residential wood framed analysis needed for 2006 IBC and ASCE 7-05

Dennis:

In my opinion the Wind Analysis based on CBC 2007 (ASCE 7-05) for low and mid rise structures are relatively easy but a lot more time consuming than the current CBC 2001.
However based on the information that I've got form Howard Smith & Bill Staelhlin down at the DSA, about two weeks ago, a new simplified Wind Design method will be available by early 2008 for California structural engineers.

Regards

Khashayar (Casey) Hemmatyar, SE
Private e: khemmatyar@hotmail.com

________________________________________
From: Dennis Wish [mailto:dennis.wish@verizon.net]
Sent: Thursday, September 13, 2007 5:42 PM
To: seaint@seaint.org
Subject: Practical Example of wind design for multi-story residential wood framed analysis needed for 2006 IBC and ASCE 7-05

I have a lot of information related to the calculations necessary to be run for compliance to the ASCE 7-05 Simplified wind design. This does not put the problem in a realistic perspective for me when comparing wind to seismic loads and the actual distribution of shear based on flexible diaphragm design. What I am looking for is an example analysis for a two story and one story building to see how wind and seismic loads are compared. I've reviewed the ASCE 7-05 Simplified wind design and it is included in the TEDDS library as well. However, while I can see how the loads are calculated for the uplift on the overhangs, leeward and windward sides and each other load condition, I am lost when trying to work a traditional example in Southern California for the design of a single or multi-story wood frame structure (less than three stories) where seismic design and wind loads can be applied using our traditional flexible design methods. For the first time in history, the seismic design is so much easier to understand than the wind loads but I suppose this makes sense in the coastal regions subject to high wind loads.

The Seismic Design manual published by ICC for the 2006 IBC volume II contains examples but I am hesitant to spend the price for the manual after being disappointed by the 97 UBC first publication of these examples with errors in place. I also received a notice that there will be a 1-1/2 hour web seminar that I can participate in for $250.00 and this is out of the questions for me. I can't possibly justify working out a solution to the wind design section (the only section that I am currently stumped on) for $250.00. The one advantage is that I can invite all of the local engineers into my office and sit around my one 19" monitor to view the webcast and spit the cost to about $40.00 per person, but I doubt if we could all see the largest screen I have and to run it on two connections according to S.K. Ghosh Associates, Inc. latest advertisement would cost an additional $250.00 or some slight adjustment. I am being facetious for a reason since $250.00 for a 90 minute seminar is something I should consider competing in. I can do it for a tenth the price and pick up almost all of the engineering community and would not mind making it a full day seminar for the profit it would return. Do you really think you can leave the seminar after 90 minutes and understand what to do? I don't think it can be done since the last time I attended the S.K. Ghosh seminar in Riverside (an hour or so away) it was a full day and came with a copy of the code. I still did not think it covered enough of the issues, but at least it was about half this price and a full day ICBO seminar.

So let me ask – has anyone started to do some sample problems (realistic designs) for light frame residential design that has a copy of your work you are willing to share with the rest of us. I want to add the wind design into my public domain MultiLat program and am delayed in getting this out to beta testers since the information I have has not reduced the confusion and frustration I am experiencing on the wind load portion.

Finally, as another pointed out, for my practice I will need the 2006 IBC (already have it), ASCE 7-05 (I've have some of the drafts), AF&PA (the most worthwhile publications when considering cost), AITC for Glu-Lam's which I believe the fourth edition is probably sufficient, NDS which may not be necessary if I stick to ASD methods, AISC (not necessary for the little steel work I do) and probably the IBC Existing Building Code because I still do some retrofit of existing buildings. Therefore, my cost is such that I don't see why most of us can avoid seminars and learn from each other off this list. If you have an office with employees, then maybe you can justify the cost of reference and seminars. If you are a one person show, then expect to be forced out of business by the cost of doing business – the competition in outsourcing is currently biting me in the butt as I am attempting to earn a living at the cost of all of the necessary references.

Thankfully, there are proprietary materials who will do some of the work for you (like MiTek's SidePlate moment frame). I heard yesterday that the City of San Diego was already planning on addendums to the IBC or state adopted code when they anticipated no addendums a few months ago. I am interested in seeing what small towns like La Quinta will do to adopt the next California code 

About 9 more years to retirement and then I will have fun spending my days writing software to give away rather than feeling the burnout of having to make a living and seeing my costs rise and profits fall.

Sorry – I am just feeling the stress of all of the latest cr*p to hit the fan..

If you have run some numbers and can share with me the practical application of a light frame single or multi-story seismic verses wind design on a residential structure using flexible diaphragm analysis, I would be greatly appreciative and will even give you a free copy of my already free software – MultiLat once it is finished 

Thanks,
Dennis

PS: If you have it in PDF format or can put in PDF format this would help. If not you can fax it to me at 1.760.564.0884

PPS: For those of you who observe, I wish you all a Happy and Healthy Jewish New Year 5768

Minimum RC column dimension, rebars

For earthquake resistant structures, RC columns are limited

to a minimum dimension of 12 in. or 305mm. , UBC 1997 1921.4.1.1 . and a minimum steel ratio of .01

For two story residential buildings with column spacing of

4.0 meters or less and height less than 7.5 meters , this results to columns with strengths in excess of what is needed to carry the required loads.

Is there any exception to this code provision that will allow an 8" or 10" RC column.?

If the factored axial load is less than 0.10Ag*Fc' will this permit me to use a 10" RC column.?

Where can I find the the code that limits stirrups and ties

to at least 3/8" ~(10mm) Dia.

Thanks in Advance.

Alex Nacionales, C.E.

Re: embedded plate

Jason,

Most of the ones I've seen use nails around the edge and bent over.

Regards,

H. Daryl Richardson

----- Original Message -----

From: Jason Atwood

To: seaint@seaint.org

Sent: Friday, September 14, 2007 8:59 AM

Subject: embedded plate

Speaking of embedded plates how do contractors set these in their forms-do they glue it with PL or something?

Jason

embedded plate

Speaking of embedded plates how do contractors set these in their forms-do they glue it with PL or something?

Jason

RE: 2007 California Building Code (5th attempt for posting!)

David:

Last week I was invited to give a presentation to a group of local structural engineers and architects. One of the topics covered, was the general information about the new 2007 CBC.

However, my answer to your question about Chapter 19 Concrete is NO.

The entire chapter 19 in CBC 2007 is now only 10 pages. For Steel, CMU, Wood, Aluminum and others, it's similar. Basically we have to use the referenced standards ( i.e. ACI-318, ACI 530, AISC, NDS) directly and include the CBC's amendments.

I have included a small portion of that presentation below: (bullets may be shown differently)

General Highlights

New CBC 2007 Part 2, Volume 2 contains:

- 2007 California Building Code (CBC) Based on IBC 2006

- 2007 California Historical Building Code; Title 24 Part 8

- 2007 California Existing Building Code; Title 24 Part 10

2007 CBC also includes:

- Chapter 31B: Public swimming pools

- Chapter 31C: Radiation (medical radiographic/X-Ray)

- Chapter 31D: Food Establishment

- Chapter 31E: Tents and Membrane Structures

- Chapter 31F: Marine Oil Terminals

Seismic Provisions Highlights

- Seismic forces based on response acceleration values determined through U.S. Geological Survey maps.

- Seismic Design Categories (SDC) A, B, C, D, E and F (A the lightest, F the most severe) replaces Seismic Zones 1 through 4.

- California mostly SDC C and up.

- For California schools & hospitals minimum SDC D

- SDC controls the design and detailing procedures

- Some Horizontal and vertical irregularities are not permitted anymore, for example:

- In Seismic Design Category E and F:

- Extreme Horizontal Torsional eccentricity,

- Extreme Soft Story

- Weak Story

- In Seismic Design Category D:

- Extreme Weak Story

- Simplified Seismic procedure is not permitted

Highlights of New Provisions Based on ASCE/SEI 7-05 for CMU & RC:

- Concrete shear-walls:

- Ordinary Plain Concert shear-walls (not permitted in SDC C, D, E & F)

- Detailed Plain concrete shear-walls (not permitted in SDC C, D, E & F)

- Ordinary reinforced concrete shear-walls (not permitted in SDC D, E & F)

- Special reinforced concrete shear-walls (permitted in all, but in SDC D, E & F height < 160 feet)

- Similar categorization applies to CMU walls:

- Pre-stressed Masonry Shear-walls is new but not allowed for California Schools & Hospitals

I hope this is helpful.
Regards

Casey (Khashayar) Hemmatyar, SE

private e: khemmatyar@hotmail.com

RE: Service Elevator Ramp

We designed the first handicap accessible marina here in San Diego. During construction of the marina, the electrical sub foreman was a wheelie. He took great joy in flying down the ramps to the floats. At the base of the ramp, the float was at right angles to the ramp and he took great pride in careening down the ramp and hooking a hard right at the base, just narrowly missing going into San Diego Bay by inches. He was a great show-off and we really enjoyed his antics.

Bob

From: Bill Polhemus [mailto:bill@polhemus.cc]
Sent: Thursday, September 13, 2007 6:00 PM
To: seaint@seaint.org
Subject: Re: Service Elevator Ramp

Daryl Richardson wrote:

Bill,

I would expect that your codes have a requirement for maximum slope for wheel chair access ramps. That might be a prudent place to start.

A refinery operator in a wheelchair.

Well, it could happen.

RE: WTC Studies-Structural Aspects

Looks like all this is something the Mythbusters (Discovery Channel) can
address. It should be an interesting show if they do.

----Original Message Follows----
From: "Alexander Bausk" <bauskas@gmail.com>
Reply-To: <seaint@seaint.org>
To: seaint@seaint.org
Subject: WTC Studies-Structural Aspects
Date: Fri, 14 Sep 2007 10:27:29 +0300

Dear Paul,

On 9/13/07, Paul Feather <PFeather@se-solutions.net> wrote:
> If I am not mistaken, even our nuclear containment facilities are only
> designed for the equivalent of an engine falling off a plane, not a
> direct kamikaze attack.

Our WWER-1000 containment building is designed to resist large airliner
crash.
As far as I know, the American and Canadian designs also consider such
impact loads.
The issue is that the analysis deals with accidental collision rather
that a malevolent act, which can result in worse consequences.
If needed, I can provide the list with links to research works that
deal with containment impacted by an airliner.

--
Alexander Bausk
CAD manager, Structural engineer at
Nuclear Engineering&Research Lab
Dnipropetrovsk, Ukraine

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Re: Service Elevator Ramp

Daryl Richardson wrote:

Bill,

I wasn't suggesting that the operator would actually be in a wheel chair; I was suggesting that the slope used for that purpose might be something you could defend if you have no other guideline. You could also try calling up your local neighborhood Occupational Health and Safety guy.

That's a good point. Thanks. Of course, my assumption is that folks here know more than any number of government employees.

Re: WTC Studies-Structural Aspects

Dick_Roberts@oxy.com wrote:
> I think you have to look at who paid for this study and was their any
> bias involved.
>

George Soros is a well-known financial backer of engineering and
scientific research. So of course there was no bias involved.

******* ****** ****** ****** ******* ****** ****** ********

RE: WTC Studies-Structural Aspects

I think you have to look at who paid for this study and was their any
bias involved.

Richard Roberts
(716)278-7147
dick_roberts@oxy.com

-----Original Message-----
From: Acharya, Suresh [mailto:Suresh.Acharya@ci.concord.ca.us]
Sent: Thursday, September 13, 2007 4:21 PM
To: 'seaint@seaint.org'
Subject: RE: WTC Studies-Structural Aspects

Questions: Did ASCE have a role in allowing the developer to skip the
then
New York building codes?

-----Original Message-----
From: Abolhassan Astaneh-Asl [mailto:astaneh@ce.berkeley.edu]
Sent: Saturday, October 13, 2007 1:14 AM
To: seaint@seaint.org
Subject: WTC Studies-Structural Aspects

Dear Friends: Yesterday, I presented the results of our 5-year studies
of structural aspects of the World Trade Center in Sibley Auditorium of
UC Berkeley. Articles in the Oakland Tribune, Contra Costa Times, and
San Jose Mercury News cover the main items of my presentation. The
articles are almost the same with minor changes.
Oakland Tribune article is at:
http://www.insidebayarea.com/oaklandtribune/localnews/ci_6870312

I welcome any and all professional and non-personal comments. A.
Astaneh-Asl

******* ****** ****** ****** ******* ****** ****** ********

RE: WTC Studies-Structural Aspects

It would be interesting to understand what NYC
building code provisions were supposedly violated in
the Design of the WTC.

Up until recently the NYC building code structurally
was out of date and an embarrassment. The code
minimum wind loads, which typically govern tall
buildings, were inadequate. The reason that there are
not more structural problems is that the engineers
designing the tall buildings had wind tunnel studies
and followed national standards.

I do not personally know what was the regulatory
climate when the WTC was constructed but my
understanding is that there is no structural plan
check in NYC. Apparently the system was so
bureaucratic and corrupt that they gave up and left it
to the design professional to self certify the design.

******* ****** ****** ****** ******* ****** ****** ********

WTC Studies-Structural Aspects

Dear Paul,

Our WWER-1000 containment building is designed to resist large airliner crash.
As far as I know, the American and Canadian designs also consider such
impact loads.
The issue is that the analysis deals with accidental collision rather
that a malevolent act, which can result in worse consequences.
If needed, I can provide the list with links to research works that
deal with containment impacted by an airliner.

--
Alexander Bausk
CAD manager, Structural engineer at
Nuclear Engineering&Research Lab
Dnipropetrovsk, Ukraine

******* ****** ****** ****** ******* ****** ****** ********

Thursday, September 13, 2007

Re: embed plate design

There is really a simple solution if you just look at
the problem differently.

The provisions in Appendix D are primarily written for
mass concrete with little or no reinforcing. If you
have enough reinforcing in your member to resist the
forces if they can be developed, you should have
little to no problem.

The secret is to make use of the strut and tie
provisions. The anchors are considered as ties that
have a head on the end embedded in the concrete. You
then define a compression strut that can be used to
engage stirrups or other bars in the reinforced
section.

******* ****** ****** ****** ******* ****** ****** ********

Re: embed plate design

I don't have a beam or column to develop the bars into -- only a round or square footing (flagpole type) x the depth required. Appendix D only allows headed anchor bolts to be embedded a maximum of 25" (D.5.2.2 ) and we usually use min 1 1/4" dia anchor rods -- tough to get a hook on that size. Lapping with a deformed bar is questionalbe at best. So I am left with increasing edge distance, increasing concrete strength, increasing number of bolts. Any of these adds up to lots of extra costs. Most of my footings now are about 6 ft + square for a 20" square base plate. If the diameter gets over 6 ft. I don't even design it for that -- just go for the square one.

Jim Persing

On 9/13/07, refugio rochin <fugeeo@gmail.com> wrote:

I think the solution to the problem lies in the development length. If you develop the anchor bolt as reinforcement ie 22 db with hook or 44 db (example) long steel, then one can develop the anchor into a column or deep beam... No? I tend to add the minimal ldh that the ACI includes for the anchor development after running through the App D to the development length.
2007/9/13, Jim Persing <omega.two.0@gmail.com>:
(Rant on) It's going to be interesting to see all of my past structures fall down (primarily gas station canopies) now that the anchor bolt connections suddenly do not work anymore. ACI 318 Appendix D is certainly an exercise in ......, well something. I'm not sure what. I now have to justify to my clients why they have to have larger footings than before with larger and longer anchor bolts (but not over 25" embedment!), higher strength bolts, higher strength concrete and special inspections that they never had to have before. My spreadsheet for anchor bolts is 4 pages long - and that's only for my special purpose anchor bolts. Between anchor bolts -- whoops, rods, and wind loads I'm glad that I'm getting close to retirement :) (Rant off)

Jim Persing, SE

On 9/12/07, Tom.Hunt@fluor.com < Tom.Hunt@fluor.com> wrote:

Christopher,

(turn on rant) Welcome to ACI 318 Appendix D. Once again a major code writing entity has slain a mighty paper dragon. It now takes half a grown tree just to design a few anchor bolts; sheeez! (turn off rant).

ACI 318 Appendix D "does" allow you to take advantage of supplemental reinforcing steel (refer to Section D.4.2.1 and it's commentary). The problem is that ACI came up with a rocket science thesis on how to design anchor bolts but provides no guidance on how to take advantage of supplementary reinforcing steel. To me this is a total let down of ACI and I have brought it up several times with Dr. Ghosh during his seminars and from his response I believe he totally agrees. There may be other or better documents but you might try finding a copy of ACI 349 Appendix B from the early to mid 1980s. Hard to believe we have to go back 25 year to find a document to help us design to a 21 Century code.

Thomas Hunt, S.E.
Fluor

"Christopher Banbury" <cbanbury@arkengineering.net >
09/11/2007 05:26 PM
Please respond to seaint

To
<seaint@seaint.org>

cc

Subject
embed plate design

I am looking for guidance on the design of steel embedment plates primarily subject to shear loads. In particular I am looking at a weld plate with anchor rods cast in a concrete column that supports a steel beam. I am unable to get the connection to work using Appendix D methodology. ACI 318 Appendix D seems to be overly conservative for calculating the capacity of the embedded anchors as it does not give significant credit for the reinforcement intersecting the failure plane.
I have had similar problems with concrete piers subject to shear and uplift loads.
Thanks in advance.

Christopher Banbury, PE
President

Ark Engineering, Inc.
PO Box 10129, Brooksville, FL 34603
22 North Broad ST, Brooksville, FL 34601
Phone: (352) 754-2424
Fax: (352) 754-2412
www.arkengineering.net

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Re: embed plate design

I think the solution to the problem lies in the development length. If you develop the anchor bolt as reinforcement ie 22 db with hook or 44 db (example) long steel, then one can develop the anchor into a column or deep beam... No? I tend to add the minimal ldh that the ACI includes for the anchor development after running through the App D to the development length.

2007/9/13, Jim Persing <omega.two.0@gmail.com>:

(Rant on) It's going to be interesting to see all of my past structures fall down (primarily gas station canopies) now that the anchor bolt connections suddenly do not work anymore. ACI 318 Appendix D is certainly an exercise in ......, well something. I'm not sure what. I now have to justify to my clients why they have to have larger footings than before with larger and longer anchor bolts (but not over 25" embedment!), higher strength bolts, higher strength concrete and special inspections that they never had to have before. My spreadsheet for anchor bolts is 4 pages long - and that's only for my special purpose anchor bolts. Between anchor bolts -- whoops, rods, and wind loads I'm glad that I'm getting close to retirement :) (Rant off)

Jim Persing, SE
On 9/12/07, Tom.Hunt@fluor.com < Tom.Hunt@fluor.com> wrote:

Christopher,

(turn on rant) Welcome to ACI 318 Appendix D. Once again a major code writing entity has slain a mighty paper dragon. It now takes half a grown tree just to design a few anchor bolts; sheeez! (turn off rant).

ACI 318 Appendix D "does" allow you to take advantage of supplemental reinforcing steel (refer to Section D.4.2.1 and it's commentary). The problem is that ACI came up with a rocket science thesis on how to design anchor bolts but provides no guidance on how to take advantage of supplementary reinforcing steel. To me this is a total let down of ACI and I have brought it up several times with Dr. Ghosh during his seminars and from his response I believe he totally agrees. There may be other or better documents but you might try finding a copy of ACI 349 Appendix B from the early to mid 1980s. Hard to believe we have to go back 25 year to find a document to help us design to a 21 Century code.

Thomas Hunt, S.E.
Fluor

"Christopher Banbury" <cbanbury@arkengineering.net >
09/11/2007 05:26 PM
Please respond to seaint

To
<seaint@seaint.org>

cc

Subject
embed plate design

I am looking for guidance on the design of steel embedment plates primarily subject to shear loads. In particular I am looking at a weld plate with anchor rods cast in a concrete column that supports a steel beam. I am unable to get the connection to work using Appendix D methodology. ACI 318 Appendix D seems to be overly conservative for calculating the capacity of the embedded anchors as it does not give significant credit for the reinforcement intersecting the failure plane.
I have had similar problems with concrete piers subject to shear and uplift loads.
Thanks in advance.

Christopher Banbury, PE
President

Ark Engineering, Inc.
PO Box 10129, Brooksville, FL 34603
22 North Broad ST, Brooksville, FL 34601
Phone: (352) 754-2424
Fax: (352) 754-2412
www.arkengineering.net

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Re: Service Elevator Ramp

Bill,

I wasn't suggesting that the operator would actually be in a wheel chair; I was suggesting that the slope used for that purpose might be something you could defend if you have no other guideline. You could also try calling up your local neighborhood Occupational Health and Safety guy.

Regards,

H. Daryl Richardson

----- Original Message -----

From: Bill Polhemus

To: seaint@seaint.org

Sent: Thursday, September 13, 2007 7:00 PM

Subject: Re: Service Elevator Ramp

Daryl Richardson wrote:

Bill,

I would expect that your codes have a requirement for maximum slope for wheel chair access ramps. That might be a prudent place to start.
A refinery operator in a wheelchair.

Well, it could happen.
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Re: Practical Example of wind design for multi-story residential wood framed analysis needed for 2006 IBC and ASCE 7-05

Dennis Wish wrote:

For the first time in history, the seismic design is so much easier to understand than the wind loads but I suppose this makes sense in the coastal regions subject to high wind loads.

"You said a mouthful!"

(Signed)
Beaumont, Texas
(H. Rita, 9/2005; H. Humberto, 9/2007)

(N.B. Dennis, go here. The wind load provisions of 7-05 are not very different at all from those of 7-02 - I know, I asked Dr. Mehta if he was going to write a guide for 7-05 and he told me "there's no need, not enough changed.")

ASCE Transactions Paper N. 1835

Dear seaint,

I'm looking for the following paper:

Transactions ASCE, Vol. 98, Paper No.1835, pp. 418-433 - Water pressures on dams during earthquakes - Harald Malcolm Westergaard.

I would really appreciate if you can send me a copy of this paper by mail.

Thanks,

Gilvan Correard

Re: Service Elevator Ramp

Daryl Richardson wrote:

Bill,

I would expect that your codes have a requirement for maximum slope for wheel chair access ramps. That might be a prudent place to start.

A refinery operator in a wheelchair.

Well, it could happen.

Re: Service Elevator Ramp

Acharya, Suresh wrote:

Max slope for a ramp is 1 in 12. Refer to ADA or your local building code for handicap accessibility requirements for buildings that will be used by public.

-Suresh Acharya, S.E.

I'm not sure that ADA would apply here. This is a hazardous operating area inside an oil refinery.

Of course, these days who knows?

Practical Example of wind design for multi-story residential wood framed analysis needed for 2006 IBC and ASCE 7-05

I have a lot of information related to the calculations necessary to be run for compliance to the ASCE 7-05 Simplified wind design. This does not put the problem in a realistic perspective for me when comparing wind to seismic loads and the actual distribution of shear based on flexible diaphragm design. What I am looking for is an example analysis for a two story and one story building to see how wind and seismic loads are compared. I’ve reviewed the ASCE 7-05 Simplified wind design and it is included in the TEDDS library as well. However, while I can see how the loads are calculated for the uplift on the overhangs, leeward and windward sides and each other load condition, I am lost when trying to work a traditional example in Southern California for the design of a single or multi-story wood frame structure (less than three stories) where seismic design and wind loads can be applied using our traditional flexible design methods. For the first time in history, the seismic design is so much easier to understand than the wind loads but I suppose this makes sense in the coastal regions subject to high wind loads.

The Seismic Design manual published by ICC for the 2006 IBC volume II contains examples but I am hesitant to spend the price for the manual after being disappointed by the 97 UBC first publication of these examples with errors in place. I also received a notice that there will be a 1-1/2 hour web seminar that I can participate in for $250.00 and this is out of the questions for me. I can’t possibly justify working out a solution to the wind design section (the only section that I am currently stumped on) for $250.00. The one advantage is that I can invite all of the local engineers into my office and sit around my one 19” monitor to view the webcast and spit the cost to about $40.00 per person, but I doubt if we could all see the largest screen I have and to run it on two connections according to S.K. Ghosh Associates, Inc. latest advertisement would cost an additional $250.00 or some slight adjustment. I am being facetious for a reason since $250.00 for a 90 minute seminar is something I should consider competing in. I can do it for a tenth the price and pick up almost all of the engineering community and would not mind making it a full day seminar for the profit it would return. Do you really think you can leave the seminar after 90 minutes and understand what to do? I don’t think it can be done since the last time I attended the S.K. Ghosh seminar in Riverside (an hour or so away) it was a full day and came with a copy of the code. I still did not think it covered enough of the issues, but at least it was about half this price and a full day ICBO seminar.

Finally, as another pointed out, for my practice I will need the 2006 IBC (already have it), ASCE 7-05 (I’ve have some of the drafts), AF&PA (the most worthwhile publications when considering cost), AITC for Glu-Lam’s which I believe the fourth edition is probably sufficient, NDS which may not be necessary if I stick to ASD methods, AISC (not necessary for the little steel work I do) and probably the IBC Existing Building Code because I still do some retrofit of existing buildings. Therefore, my cost is such that I don’t see why most of us can avoid seminars and learn from each other off this list. If you have an office with employees, then maybe you can justify the cost of reference and seminars. If you are a one person show, then expect to be forced out of business by the cost of doing business – the competition in outsourcing is currently biting me in the butt as I am attempting to earn a living at the cost of all of the necessary references.

Thankfully, there are proprietary materials who will do some of the work for you (like MiTek’s SidePlate moment frame). I heard yesterday that the City of San Diego was already planning on addendums to the IBC or state adopted code when they anticipated no addendums a few months ago. I am interested in seeing what small towns like La Quinta will do to adopt the next California code J

Sorry – I am just feeling the stress of all of the latest cr*p to hit the fan..

Thanks,

Dennis

PS: If you have it in PDF format or can put in PDF format this would help. If not you can fax it to me at 1.760.564.0884

PPS: For those of you who observe, I wish you all a Happy and Healthy Jewish New Year 5768

Re: Retaining Wall Questions

Jordan,

Let me suggest a cross-section for your consideration. For the front (call it the right, for discussion purposes) face of the trial cross-section use the front faces of the upper and lower retaining walls as you would like them to appear; for the back (make it the left) face of the trial cross section use a vertical line at the farthest back (left) of the following three points: 1) the top back of the upper retaining wall; 2) the back (left) edge of the footing of the upper wall; or 3) the back (left) edge of the footing of the lower wall. If this trial cross-section will not work as a gravity retaining wall then you must change the section or use tie backs. If the section does work you should be able to design both the components and their interconnection.

With retaining walls, over designing is probably a good idea. The cost of over designing (call it the insurance premium) is relatively trivial; the cost of failure is VERY substantial.

Regards,

H. Daryl Richardson

----- Original Message -----

From: Wesley Werner

To: seaint@seaint.org

Sent: Thursday, September 13, 2007 3:56 PM

Subject: RE: Retaining Wall Questions

Jordan,

This doesn't exactly answer your question, but at a seminar I attended on segmental retaining walls the presenter said that you had to design the walls as one wall unless the horizontal distance between them was at least twice the height of the lower wall. He didn't give us the theory behind the rule-of-thumb, though. Also, he said that you needed to check the global stability of the whole tiered system. He said that he had seen at least one case where a whole hillside gave way behind the walls rather than the walls themselves giving way.

Wesley C. Werner

-----Original Message-----
From: Jordan Denio [mailto:jordan@AshleyVance.com]
Sent: Thursday, September 13, 2007 5:28 PM
To: seaint@seaint.org
Subject: RE: Retaining Wall Questions

I don't have the answer for that as I'm looking for opinions/advice to determine what the size and configuration of the walls should be. If it helps the discussion, let's say the wall upper wall is set back 5' from the lower wall and all the sliding is resisted by a key. The lower wall retains ~8'-0" and the upper wall retains ~4'-0".

From: Jnapd@aol.com [mailto:Jnapd@aol.com]
Sent: Thursday, September 13, 2007 9:10 AM
To: seaint@seaint.org
Subject: Re: Retaining Wall Questions

Jordan

How far apart are the walls...vertically and Horizontally

See what's new at AOL.com and Make AOL Your Homepage.

RE: Heavy or light storage load

Bob,

That's more what I was looking for. I wasn't quite sure about auto part weights. It has been a long time since I did any auto work, but the last thing I did was replace a starter in my '75 Ford F100 about 10 years ago. I don't remember well enough what the weight might have been, but I remember it being somewhat heavy. I just wanted to know if someone had some numbers to apply to this situation. Sounds like the 250 psf is the more likely.

Jeff Hedman

Need some advice for a rafter tie splice

Please respond to me privately at dennis.wish@verizon.net unless you feel
that this e-mail has merit to anyone else who may benefit from it.

Problem:
A client has reframed the roof of a garage without a permit and must bring
the structure up to code through engineering. The method of construction
contains a few non-conventional connections that are creating a problem in
my retrofit design and detail.

1. The dimensions of the garage are 24-feet wide and 21-feet deep. The
structure is existing and while it probably does not comply with current
code for lateral design, the building department has only asked the owner to
have an engineer justify or retrofit the portion of the garage that was
modified - the roof framing. The weight of the new dead load materials is
actually less than the original design and the live load will not have
changed. Because the total load is less than or equal to what the original
roof was designed for, the building official is not asking that lateral
design be considered as the performance will not have changed other than in
the gravity load design.

2. The original roof was a flat roof and this was removed and replaced with
2x6 rafters @ 16" o.c. sloped to 1.5" per foot. The contractor installed a
2x12 DF#2 ridge board to span the 21-feet and believed this to be adequate
as a ridge beam. He did not install any ceiling joists/ties. Therefore there
is an outward thrust on the exterior walls and the potential for the 2x12
ridge to deflect as much as 5-inches under the weight of the new foam roof.

3. The 2x6 DF#2 rafters are extended over the bearing walls to create a
21-inch eave. The 2x6's were end nailed through the 2x12 on one side. The
2x6 on the other side of the 2x12 was aligned end to end with the other 2x6
and toe-nailed through the 2x12 into the sides of the 2x6.

4. The 2x6 was not seat cut on the double 2x4 bearing wall and sits at a
7+degree angle (1.5:12). Solid blocking was installed by at the same angle
of the 2x6 rafters and set with the inside edge of the 2x6 block at the
outside edge of the 2x4 stud wall. The 2x6 rafters were connected to the 2x4
double top plate using a Simpson A34 or A35 and the blocking was toe-nailed
to the rafters.

5. The roof was sheathed with 19/32" APA rated exterior glue plywood and
nailed to the rafters at 6:6 nailing using 8d sinkers as I suspect rather
than common nails.

6. The panels were not blocked but the pattern was laid properly with the
ends of panels on the rafters and each row of panels staggered. Field
Nailing was at 6" o.c. rather than 12" required.

Intended Corrections:

A. The weight of the foam applied over the 19/32" plwd is only 0.25-psf for
1-inch of foam and only a maximum of 1-inch thick foam can be applied before
it must be stripped off and replaced. The panel nailing is considered
adequate.

B. The slope of the roof with the 2x12 DF#2 Ridge board still allows me
clearance to install 2x6 DF#2 rafter ties. The problem is the tension in the
ties will require a rafter tie at every rafter. Based on a live load of 20
psf and a minimum dead load of 10-psf (much higher than actual) the reaction
at the bearing walls is 556-lbs and because of the low slope of the roof,
the horizontal load that causes thrust in the exterior walls is 4436.88
pounds.

C. This requires the splice of (2) 2x6's to make up the 24-feet across the
garage. I can face nail the tie to the rafters with 16d x 3" long nails and
following the AITC requirement for single shear with Z=133 pounds per nail
and a permanent wind or seismic load factor Cd of 1.6 and because we are
here in the desert where over half of the year the temperature can be easily
over 110 degrees in an unconditioned space, the Ct factor is 0.8. Therefore
I would need 26 nails (Hankinson's formula was not considered for nailed
connections according to my interpretation of the AITC Table 7.52 for Common
Wire Nails). Simpson makes a 16d nail that is 3-inches long so embedment is
not an issue as much as shank size since 10d nails would require almost
twice the number at each connection.

D. HERE IS THE KICKER OF THE PROBLEM - THE RAFTERS ARE ALL ALIGNED ON EACH
SIDE OF THE 2X12. THE FRAMER IS NOT ADVERSE TO ADDING A SIMPSON LSU26
(SLOPED BOTTOM FLUSH JOIST HANGER) ON THE SIDE WHERE THE NAILS ARE
TOE-NAILED TO THE 2X12, BUT I MUST SPLICE THE 2X6 RAFTER TIES END TO END AND
THE LOAD IN TENSION/COMPRESSION IS 4,436.88 POUNDS. I DON'T WANT TO SISTER
ON A PARTIAL 2X6 TO ONE SIDE SINCE IT IS LATERALLY UNSUPPORTED AND I WORRY
ABOUT TWIST IN THE TIE AT THE CONNECTION WHETHER I USE NAILS OR SCREWS - SO
I WANT TO USE A BOLTED CONNECTION BUT EITHER NEED TO FABRICATE A PLATE FOR
EVERY SPLICE OR USE A SIMPSON STRAP CONNECTION (MY THOUGHT IS A STRAP ON
EACH SIDE AT TOP AND BOTTOM OR FOUR STRAPS TOTAL PER SPLICE) TO PICK UP THE
4436 POUND LOAD.

Now I need some advice. I will come back in and connect the blocking to the
top plate of the wall using a Simpson RBC clip bent to the angle of the
block to top plate to restore any shear transfer (the plywood is boundary
nailed to the blocking).

Can anyone help me with a reasonable solution to the rafter tie splice
problem? I spoke to the framer and he has no trouble with the nailing of the
ties to the rafters using a palm nailer. He can do it quickly and easily. He
also has no problem with a splice or adding the hanger on one side since
this would repair the problem he caused much less expensively than tearing
it all out.

NOTE: I SUGGESTED AT CONTINUOUS TIMBERSTRAND 1-1/2" MEMBER BUT HE FEELS THAT
THE COST OF THE LUMBER AND SPECIAL ORDER WOULD BE MUCH MORE EXPENSIVE FOR
HIM THAN TO MAKE THE SPLICES IN THE FILED.

I am late on getting this out because of the small issues in his method of
construction that created a much more difficult means of installing the ties
as well as the high tension load to the tie due to the very low slope of the
roof. I can use some advice but I need it quickly and will check incoming
e-mail much faster than I do the list digest mode. SO PLEASE E-MAIL ME
DIRECTLY AT DENNIS.WISH@VERIZON.NET.

Thank you,
Dennis

P.S. THE ENDS OF THE GARAGE ARE GABLED ENDS AND ARE NOT A PROBLEM. THE
FRAMER THOUGHT THE 2X12 SPANNING 21-FEET WOULD BE SUFFICIENT AS A BEARING
RIDGE UNTIL I WAS CALLED IN TO EXPLAIN THE DIFFERENCE BETWEEN A RIDGE BEAM
AND A RIDGE BOARD. ALSO I TRIED TO RUN THIS WITH THE TIES AT 4'-0" BUT AS
YOU CAN EXPECT, THE TENSION/COMPRESSION IN THE SPLICE AN AT THE RAFTERS WAS
OVER 12,000 POUNDS.

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