Structural Engineering: 2007-08-05

Saturday, August 11, 2007

Re: Expansion / control joint

Doug, thanks for that link, the info was very good.

On 8/11/07, William.Sherman@ch2m.com <William.Sherman@ch2m.com > wrote:

It is the engineer's responsibility to provide direction regarding use of movement joints. The mason should not be providing joints that interrupt structural continuity without direction from the engineer.

Bill Sherman

CH2M HILL / DEN

720-286-2792

From: erik gibbs [mailto:erik.gibbs@gmail.com]
Sent: Friday, August 10, 2007 3:29 PM
To: seaint@seaint.org
Subject: Expansion / control joint

I recieved a plan check on a cmu retaining wall that I designed asking to provide a detail for rebar placement for expansion joints. I realize that on long walls expansion joints should be used, but as far as specifying them on structural drawings, shouldn't that be the job of the mason? I looked in the UBC and I couldn't find any info about this. Has anyone come across this before?

Thanks

Erik Gibbs

It is the engineer's responsibility to provide direction regarding use of movement joints. The mason should not be providing joints that interrupt structural continuity without direction from the engineer.

Bill Sherman

CH2M HILL / DEN

720-286-2792

From: erik gibbs [mailto:erik.gibbs@gmail.com]
Sent: Friday, August 10, 2007 3:29 PM
To: seaint@seaint.org
Subject: Expansion / control joint

I recieved a plan check on a cmu retaining wall that I designed asking to provide a detail for rebar placement for expansion joints. I realize that on long walls expansion joints should be used, but as far as specifying them on structural drawings, shouldn't that be the job of the mason? I looked in the UBC and I couldn't find any info about this. Has anyone come across this before?

Thanks

Erik Gibbs

Re: Notched Beam

sounds like you may simply add a top flange hanger to the side of the beam, depending on the load a HUCTF (or similar) will fit under the beam and onto the top flange. I dont have my catalog in front of me, verify no face nailing required on the huctf. Depending on the load distribution, the top and bottom of the beam will both be in bearing, and depending on the notch length from the face of the hanger, all the load may be justified as taken by the hanger.

If you cant locate an appropriate simpson product, welding a custom hanger to the beam sounds like it would be more cost effective than replacing the beam.

IRV FRUCHTMAN <ifaeng@yahoo.com> wrote:

Dear Fellow Engineers,
On an ongoing renovation project I observed the
following framing detail:

A 12" deep PSL beam is supported by a W12-72 I-beam,
with the top of the PSL some 7" above the top of the
I-beam. The builder notched the end of the PSL to
"fit" past the top flange (plus 2by nailer) and extend
to the web of the I-beam. The notch is about 7" from
the top of the PSL and 2" wide and 6" long. The bottom
of the PSL is blocked to the bottom flange.

I calculate that PSL shear stress is about twice the
allowed by assuming the notch extends to the bottom of
the beam.

Before I advise the owner to replace the PSLs (there
are 4) am I missing something or is there is a way to
reinforce them?

TIA,
Irv

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Re: Notched Beam

Irv,
I gather from what you have descibed that the notch effect is really on
the top, as you are saying that the "bottom of the PSL is blocked to the
bottom flange". I can't comment on what your codes require, but the
Canadian wood code is mostly concerned with notches in the bottom
portion of the beam or joist. If you have bottom bearing, then it may
not be a big concern, but perhaps some other lister familiar with the US
requirements can comment.
About a year ago, I was requested to look at a building approx 40 years
old that had wood roof trusses with the bottom chord on one end extended
approx 8"-12" so that it acted as a short cantilever. Some of these had
been notched on the bottom and almost all of them had cracks starting at
the notch and extending past the top chord-bottom chord intersection. I
told the Owner they were a real problem and he said leave it with
him-never heard from him. I think I would probably have had the
pertinent trusses jacked up and screws inserted in to the split pieces
from the bottom, and also I would have nailed on new lumber on each side
of the bottom chord.
Gary

IRV FRUCHTMAN wrote:
> Dear Fellow Engineers,
> On an ongoing renovation project I observed the
> following framing detail:
>
> A 12" deep PSL beam is supported by a W12-72 I-beam,
> with the top of the PSL some 7" above the top of the
> I-beam. The builder notched the end of the PSL to
> "fit" past the top flange (plus 2by nailer) and extend
> to the web of the I-beam. The notch is about 7" from
> the top of the PSL and 2" wide and 6" long. The bottom
> of the PSL is blocked to the bottom flange.
>
> I calculate that PSL shear stress is about twice the
> allowed by assuming the notch extends to the bottom of
> the beam.
>
> Before I advise the owner to replace the PSLs (there
> are 4) am I missing something or is there is a way to
> reinforce them?
>
> TIA,
> Irv
>
>
>
>
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Notched Beam

Dear Fellow Engineers,
On an ongoing renovation project I observed the
following framing detail:

A 12" deep PSL beam is supported by a W12-72 I-beam,
with the top of the PSL some 7" above the top of the
I-beam. The builder notched the end of the PSL to
"fit" past the top flange (plus 2by nailer) and extend
to the web of the I-beam. The notch is about 7" from
the top of the PSL and 2" wide and 6" long. The bottom
of the PSL is blocked to the bottom flange.

I calculate that PSL shear stress is about twice the
allowed by assuming the notch extends to the bottom of
the beam.

Before I advise the owner to replace the PSLs (there
are 4) am I missing something or is there is a way to
reinforce them?

TIA,
Irv

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Friday, August 10, 2007

RE: Expansion / control joint

We will often either show expansion joint locations on the plans or in a typical detail that calls out a specific spacing. The Concrete Masonry Association of California and Nevada provides excellent guidelines for the location and spacing of joints in their “Design of Reinforced Masonry Structures” and they also have an issue of “Masonry Chronicles” dealing with the subject. The “Masonry Chronicles” issue can be found online here: http://www.cmacn.org/publications/chronicles/winter02-03/winter02-03_index.htm.

HTH,

Doug Mayer, SE

Structural Engineer

TaylorTeter

Partnership

7535 North Palm Ave., Suite 201

Fresno, CA 93711

(559) 437-0887 Ph.

(559) 438-7554 Fax

doug.mayer@taylorteter.com

From: erik gibbs [mailto:erik.gibbs@gmail.com]
Sent: Friday, August 10, 2007 2:29 PM
To: seaint@seaint.org
Subject: Expansion / control joint

Thanks

Erik Gibbs

Expansion / control joint

Thanks

Erik Gibbs

Re: Access Hatch, H20 Wheel Loadings for off street and on street

For conditions other than parking, off street loading does not account for impact, and street (bridge) loading does.

----- Original Message -----
From: "Lakhani, Sid"
To: seaint@seaint.org
Subject: Access Hatch, H20 Wheel Loadings for off street and on street
Date: Fri, 10 Aug 2007 09:35:28 -0700

What is the difference between off street wheel loading and on street loadings?

Bilco has access hatch available for off street H20 wheel loading. Can I buy access hatch for on street hatch subjected to traffic loading.

Thanks!

Sid Lakhani

Bart Needham, SE Principal, nbse associates, inc. Office 206-780-6822 Office 805-452-8152 Fax    206-780-6683 Fax    208-693-3667 Mobile 206-300-2346  Office locations: 629 State Street #230 Santa Barbara, CA  93101  205 Fairview Lane Suite 100 Paso Robles, CA  93446  365 Ericksen Ave. NE Suite 328 Bainbridge Island, WA  98110  Mail and Deliveries: 321 High School Rd. NE Suite D-3 PMB 216 Bainbridge Island, WA  98110

RE: Access Hatch, H20 Wheel Loadings for off street and on street

To my information off street wheel loading for hatch is considered with occasional h20 loading design while street loading is regular H20 loading based on full traffic.

They should have both

Check with Bilco they explains you.

Best Regards,

Sanjay Kumar Verma, P.E.

-----Original Message-----
From: Lakhani, Sid [mailto:slakhani@ebmud.com]
Sent: Friday, August 10, 2007 9:35 AM
To: seaint@seaint.org
Cc: Lakhani, Sid
Subject: Access Hatch, H20 Wheel Loadings for off street and on street

What is the difference between off street wheel loading and on street loadings?

Bilco has access hatch available for off street H20 wheel loading. Can I buy access hatch for on street hatch subjected to traffic loading.

Thanks!

Sid Lakhani

Access Hatch, H20 Wheel Loadings for off street and on street

What is the difference between off street wheel loading and on street loadings?

Bilco has access hatch available for off street H20 wheel loading. Can I buy access hatch for on street hatch subjected to traffic loading.

Thanks!

Sid Lakhani

Thursday, August 9, 2007

RE: dual system

That is not a dual system. You will need to use the R value for light framed shear walls.

To the best of my understanding you would need either a braced frame or a concrete shear wall _in line_ with the frame to consider it a dual system. Lateral separation is not acceptable except maybe a few feet

Mark E. Deardorff, SE
R & S Tavares Associates, Inc
9815 Carroll Canyon Road
Suite 206
San Diego, CA 92131
Phone: 858-444-3344
Phone: 209-863-8928
Cell: 209-765-5592
mark@rstavares.com
www.rstavares.com

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From: DA ENGINEERING [mailto:dnae@cox.net]
Sent: Thursday, August 09, 2007 12:41 PM
To: seaint@seaint.org
Subject: dual system

Hi

just wonder what dual system?

is that means if I have residential house one steel frame at front

and shear wall 20 apart in the same direction

consider a dual system and I have to design the frame

for 25% of base shear

section cbc 1629.6.5

it a house , flexible diaphragm

Thanks

Dave A.

dual system

just wonder what dual system?

is that means if I have residential house one steel frame at front

and shear wall 20 apart in the same direction

consider a dual system and I have to design the frame

for 25% of base shear

section cbc 1629.6.5

it a house , flexible diaphragm

Thanks

Dave A.

RE: Deflection Limits for Studs Backing Brick Veneer

Chapter 16 of the Florida Building Code and IBC both allow wind loads to be
taken as 0.7 times the "component and cladding" wind loads for the purposes
of determining deflection limits.

Our firm takes the approach of meeting the L/600 deflection criteria
commonly specified but we do so by checking the deflection limit using the
components and cladding wind load reduced by 30%. Design for moment, shear,
etc. is checked using the full wind load.

This seems to be in keeping with the idea of "service level" wind
recommended by the Brick Institute.

Sean Martin
David H. Melvin, Inc.
Tallahassee, Florida

-----Original Message-----
From: Harold Sprague [mailto:spraguehope@hotmail.com]
Sent: Wednesday, August 08, 2007 12:48 PM
To: seaint@seaint.org
Subject: RE: Deflection Limits for Studs Backing Brick Veneer

Bill,
I agree that L/600 is too stringent for out-of-plane bending for a
serviceability issue. The Canadadian research "Technics Steel Stud / Brick
Veneer Walls", by Trestain and Rousseau is one of the best studies and drew
from the McMaster University studies. The McMaster studies actually
constructed veneer stud walls and tested with wind pressure and simulated
rain.

The result was that there was no increased system vulnerability due to
excessive leakage from the flexural cracking. The L/720, 600, 360 or
whatever does not elmininate flexural cracking. The deflection limit is
intended to reduce the flexural cracking size. But as the McMaster study
indicated, the size of the flexural cracking did not increase the system
vulnerability.

What did have a more significant effect on the system were the elements to
control and manage the moisture that enters through the brick from rain and
dew point and provide corrosion resistance. The Technics article did
recommend L/720 for the full wind load, but (as stated earlier) actually
provided evidence that the crack width was not an issue for system
performance.

A case can be made to use L/400 for the 50 year design wind (inferring the
L/600 for a 10 year service). I also suggest a look over the architect's
shoulder to see if the system is properly accounting for water management
and corrosion resistance.

Regards,
Harold Sprague

>From: <William.Sherman@CH2M.com>
>Reply-To: <seaint@seaint.org>
>To: <seaint@seaint.org>
>Subject: RE: Deflection Limits for Studs Backing Brick Veneer
>Date: Wed, 8 Aug 2007 07:12:18 -0600
>
>I feel that a reference to "service level wind loads" without a
>qualifier means code based wind loads without load factors applied.
>Thus, it would mean a 50-year wind load as written.
>
>But I do agree that the issue of "serviceability" is much more
>subjective. I think that a deflection limit of L/720 makes more sense
>for vertical deflection of lintels than for out-of-place deflection of
>masonry walls, due to greater wall flexibility in the out-of-plane
>direction. I would prefer to see the deflection limit defined for full
>code level, "service level wind loads", than define it for a lesser wind
>frequency, even if the lesser wind frequency is part of the basis for
>the defined limit. This just keeps requirements more "user friendly".
>
>Ultimately, I tend to feel that L/600 is too stringent a limitation for
>out-of-plane deflection.
>
>
>Bill Sherman
>CH2M HILL / DEN
>720-286-2792
>
>-----Original Message-----
>From: Harold Sprague [mailto:spraguehope@hotmail.com]
>Sent: Tuesday, August 07, 2007 9:40 AM
>To: seaint@seaint.org
>Subject: Deflection Limits for Studs Backing Brick Veneer
>
>There has been some good discussion on the maximum deflections of studs
>that back up brick veneer. There have been many good papers on the
>topic.
>Promulgated deflection limits include L/360 (steel stud mfgrs.), L/600
>(BIA), and L/720 (Canadian Research).
>
>Interestingly, the BIA guidance (TEK Note 28 B) limits the lateral
>deflection of the stud to L/600 for "service" wind loads. Per BIA 28B,
>"Therefore, to obtain sufficient backing stiffness, the allowable
>out-of-plane deflection of the studs due to service level loads should
>be restricted to L/600." But BIA does not define "service level loads".
>
>For wind the IBC and ASCE 7 have us calculate the variable "p" that is
>defined as the "design" wind pressure and is the 50 year Mean Recurrence
>Interval (MRI). Serviceability is discussed in the ASCE 7 Section
>C6.5.5 and in the AISC Design Guide 3. The general consensus of the
>AISC is that service level winds are 10 year MRI winds and are about 75%
>of the pressure calculated from "design" 50 year MRI winds.
>
>If the above logic is considered valid, the L/600 BIA limit at a
>"service"
>10 year MRI wind would be about the same as a L/400 at a 50 year MRI
>"design" wind load.
>
>I know it is conservative to use the 50 year MRI for the L/600, but it
>also increases the cost. I would welcome discussion and any performance
>studies on systems constructed.
>
>Building codes focus on life safety. This is a serviceability issue.
>
>Regards,
>Harold Sprague
>
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Wednesday, August 8, 2007

RE: Deflection Limits for Studs Backing Brick Veneer

Regards,
Harold Sprague

http://www.seaint.org/sealist1.asp
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RE: Deflection Limits for Studs Backing Brick Veneer

I feel that a reference to "service level wind loads" without a
qualifier means code based wind loads without load factors applied.
Thus, it would mean a 50-year wind load as written.

But I do agree that the issue of "serviceability" is much more
subjective. I think that a deflection limit of L/720 makes more sense
for vertical deflection of lintels than for out-of-place deflection of
masonry walls, due to greater wall flexibility in the out-of-plane
direction. I would prefer to see the deflection limit defined for full
code level, "service level wind loads", than define it for a lesser wind
frequency, even if the lesser wind frequency is part of the basis for
the defined limit. This just keeps requirements more "user friendly".

Ultimately, I tend to feel that L/600 is too stringent a limitation for
out-of-plane deflection.

Bill Sherman
CH2M HILL / DEN
720-286-2792

-----Original Message-----
From: Harold Sprague [mailto:spraguehope@hotmail.com]
Sent: Tuesday, August 07, 2007 9:40 AM
To: seaint@seaint.org
Subject: Deflection Limits for Studs Backing Brick Veneer

There has been some good discussion on the maximum deflections of studs
that back up brick veneer. There have been many good papers on the
topic.
Promulgated deflection limits include L/360 (steel stud mfgrs.), L/600
(BIA), and L/720 (Canadian Research).

Interestingly, the BIA guidance (TEK Note 28 B) limits the lateral
deflection of the stud to L/600 for "service" wind loads. Per BIA 28B,
"Therefore, to obtain sufficient backing stiffness, the allowable
out-of-plane deflection of the studs due to service level loads should
be restricted to L/600." But BIA does not define "service level loads".

For wind the IBC and ASCE 7 have us calculate the variable "p" that is
defined as the "design" wind pressure and is the 50 year Mean Recurrence
Interval (MRI). Serviceability is discussed in the ASCE 7 Section
C6.5.5 and in the AISC Design Guide 3. The general consensus of the
AISC is that service level winds are 10 year MRI winds and are about 75%
of the pressure calculated from "design" 50 year MRI winds.

If the above logic is considered valid, the L/600 BIA limit at a
"service"
10 year MRI wind would be about the same as a L/400 at a 50 year MRI
"design" wind load.

I know it is conservative to use the 50 year MRI for the L/600, but it
also increases the cost. I would welcome discussion and any performance
studies on systems constructed.

Building codes focus on life safety. This is a serviceability issue.

Regards,
Harold Sprague

Re: Deflection of Wood Studs for Brick

Well, to be a pain, it's because controlling the slope at the support
under a uniform load, you control the rate of change of the slope, which
is somewhat analogous to the imposed strain at the mortar joints. Don't
feel bad, mechanicals do this too when designing for serviceability in
composite circuit boards - we just do it in two dimensions instead of one.

There are two issues - the first is that above, the second is that for
many years brick veneer was not held to such tolerances, and the
resulting stiffness of the backup is far greater than people "expect,"
especially for wood backup. This is a little easier to get away with in
steel, because you have several gauges and flange configurations in CFS
which allow you to "hide" the increased I. In wood, going from the L/72
I mentioned as allowable in the IRC (or about L/90 in the IBC, table
2308.9.1) to L/600 at design loads means going from a 2x4 stud to a 2x8
stud in a "normal" wall. Even with data, it can be difficult to argue
that someone should double or triple the cost of a wall, and have it be
no "safer" - just less likely to leak or show cracks. This is especially
true when we design for 50 year loads, and the statute of repose is a
small fraction of that. (Difficult to argue about, not difficult to come
up with an answer on paper - the engineering calcs are always easier
than telling someone they cannot afford to build their building, or
build it for a price which provides a competitive lease rate in the market)

I suspect the real answer you're going to get to your question is that
(a) it correlates well to the limit for damage - say, within 10 or 15% -
and (b) the additional time and effort it takes to calculate the exact
answer is generally not justifiable given (a). Some engineers, given
the L/600 number as a guideline will design to it, then look at the
answer they get and if the deflection calcs to L/500 or better for the
"worst case" call it good and go on. Until I start modeling the
stiffness of the sheathing on both sides, the semi-fixed rotation spring
constant for the track at the top and bottom, and account for the short
and long term friction between the nails and the plates, I'm going to
fall squarely in the "some engineers" camp above.

Jordan

Mark Gilligan wrote:
> I believe that part of the problem with establishing
> deflection criteria for masonry on metal studs has to
> do with the way we specify deflection criteria.
>
> When I see deflection criteria specified in terms of
> L/X I wonder why they are limiting the slope of the
> beam at the support. If you look at the equations for
> deflection and compare it to deflection criteria in
> the form of L/X it is clear that all beams with L/360
> have the same slope at the support.
>
> If you are interested in controlling the natural
> frequency you should specify deflection criteria in
> inches. Similarly if you are interested in limiting
> the curvature in your beam you should specify
> deflection criteria in terms of L^2/X.
>
> What we need to do is to decide what we need to limit
> and then specify criteria that controls that
> parameter. Until we find a rational way to specify
> criteria we will have difficulties whenever our spans
> are significantly different from those tested in
> establishing the criteria.
>
> The next time you are checking beam deflections ask
> yourself why you are controlling the slope of the beam
> at the support.
>
> Mark Gilligan
>
> ttp://www.seaint.org

> ******* ****** ****** ****** ******* ****** ****** ********
>
>

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Tuesday, August 7, 2007

Re: Deflection of Wood Studs for Brick

I believe that part of the problem with establishing
deflection criteria for masonry on metal studs has to
do with the way we specify deflection criteria.

When I see deflection criteria specified in terms of
L/X I wonder why they are limiting the slope of the
beam at the support. If you look at the equations for
deflection and compare it to deflection criteria in
the form of L/X it is clear that all beams with L/360
have the same slope at the support.

If you are interested in controlling the natural
frequency you should specify deflection criteria in
inches. Similarly if you are interested in limiting
the curvature in your beam you should specify
deflection criteria in terms of L^2/X.

What we need to do is to decide what we need to limit
and then specify criteria that controls that
parameter. Until we find a rational way to specify
criteria we will have difficulties whenever our spans
are significantly different from those tested in
establishing the criteria.

The next time you are checking beam deflections ask
yourself why you are controlling the slope of the beam
at the support.

Mark Gilligan

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RE: Prequalified Welds in the AISC Manual, 13th Edition

AISC Specification references AWS D1.1-04.

AWS D1.1 is updated frequently (i.e. 2005, 2006) so the engineer should specify the version that applied to the contract documents.

Bill

From: Garner, Robert [mailto:rgarner@moffattnichol.com]
Sent: Tuesday, August 07, 2007 9:12 AM
To: Garner, Robert; seaint@seaint.org
Subject: RE: Prequalified Welds in the AISC Manual, 13th Edition

AISC just got back to me - this was corrected in the second printing of the Manual.

Compliments to AISC on their prompt answer.

Bob Garner, S.E.

From: Garner, Robert [mailto:rgarner@moffattnichol.com]
Sent: Tuesday, August 07, 2007 9:42 AM
To: seaint@seaint.org
Subject: Prequalified Welds in the AISC Manual, 13th Edition

Good morning all,

I just sent a notice in to AISC noting that Table 8-2 identifies applicable Notes by numbers, however, on Page 8-34, the Notes are listed alphabetically. Also, the AISC Manual does not seem to be updated to AWS D1.1-06. In fact, I can't tell which version of AWS D1.1 is the basis of the AISC Manual. That would be nice to know when you're trying to coordinate contract documents.

Bob Garner, S.E.

R. Garner

Moffatt & Nichol

Tel.: (619) 220-6050

Fax.: (619) 220-6055

e-mail: rgarner@moffattnichol.com

The information contained in the e-Mail, including any accompanying documents or attachments, is from Moffatt & Nichol and is intended only for the use of the individual or entity named above, and is privileged and confidential. If you are not the intended recipient, be aware that any disclosure, dissemination, distribution, copying or use of the contents of this message is strictly prohibited. If you received this message in error, please notify us.

RE: Prequalified Welds in the AISC Manual, 13th Edition

AISC just got back to me - this was corrected in the second printing of the Manual.

Compliments to AISC on their prompt answer.

Bob Garner, S.E.

From: Garner, Robert [mailto:rgarner@moffattnichol.com]
Sent: Tuesday, August 07, 2007 9:42 AM
To: seaint@seaint.org
Subject: Prequalified Welds in the AISC Manual, 13th Edition

Good morning all,

Bob Garner, S.E.

R. Garner

Moffatt & Nichol

Tel.: (619) 220-6050

Fax.: (619) 220-6055

e-mail: rgarner@moffattnichol.com

Prequalified Welds in the AISC Manual, 13th Edition

Good morning all,

Bob Garner, S.E.

R. Garner

Moffatt & Nichol

Tel.: (619) 220-6050

Fax.: (619) 220-6055

e-mail: rgarner@moffattnichol.com

Deflection Limits for Studs Backing Brick Veneer

There has been some good discussion on the maximum deflections of studs that
back up brick veneer. There have been many good papers on the topic.
Promulgated deflection limits include L/360 (steel stud mfgrs.), L/600
(BIA), and L/720 (Canadian Research).

Interestingly, the BIA guidance (TEK Note 28 B) limits the lateral
deflection of the stud to L/600 for "service" wind loads. Per BIA 28B,
"Therefore, to obtain sufficient backing stiffness, the allowable
out-of-plane deflection of the studs due to service level loads should be
restricted to L/600." But BIA does not define "service level loads".

For wind the IBC and ASCE 7 have us calculate the variable "p" that is
defined as the "design" wind pressure and is the 50 year Mean Recurrence
Interval (MRI). Serviceability is discussed in the ASCE 7 Section C6.5.5
and in the AISC Design Guide 3. The general consensus of the AISC is that
service level winds are 10 year MRI winds and are about 75% of the pressure
calculated from "design" 50 year MRI winds.

If the above logic is considered valid, the L/600 BIA limit at a "service"
10 year MRI wind would be about the same as a L/400 at a 50 year MRI
"design" wind load.

I know it is conservative to use the 50 year MRI for the L/600, but it also
increases the cost. I would welcome discussion and any performance studies
on systems constructed.

Building codes focus on life safety. This is a serviceability issue.

Regards,
Harold Sprague

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Re: Glu-Lam beam with Steel Channel calc

In a message dated 8/7/07 5:13:27 AM, t.w.allen@cox.net writes:

Jack the (E) GLB to take out all of the deflection, attach the new steel section to the GLB (using the GLB only to transfer the load) then remove the shoring.

I recommend that you jack it up additionally to camber it for at least the dead load deflection of the new steel beam, *before* connecting the steel to the wood. On a similar recent project I had them fasten the steel to the wood in only about the center one-third of the span, then jack up the ends of the steel before drilling through the wood for the connections. I never heard anything back so I assume it was workable and successful. Just a suggestion.

Ralph

Ralph
Ralph Hueston Kratz

Rhkratzse@aol.com

510-236-6668
Fax 510-215-2430

724 McLaughlin Street
Richmond CA 94805-1402 USA

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RE: 2007 California Building Code

There is no difference between the IBC and CBC if you are designing standard occupancy structures. The only amendments California has made to the model code is for special occupancies such as schools and hospitals.

Ben Yousefi, SE
Santa Monica, CA

________________________________

From: John Atilano [mailto:jatilano@laneengineers.com]
Sent: Mon 8/6/2007 3:30 PM
To: seaint@seaint.org
Subject: 2007 California Building Code

Has anyone had a chance to look at the 2007 California Building Code? How does it treat a regular commercial building? Does it simply refer to the ASCE 7-05 or does it make a lot of modifications? I figured for school or hospital projects there will be many modifications, but I was more curious about a simple commercial building. I'm trying to determine how many copies to buy for our office.

Thank you,

John Atilano, P.E.

RE: Glu-Lam beam with Steel Channel calc

Some years ago, prior to the easy availability of LVL's, PSL's, etc, Contractors could buy off-the-shelf sectios of steel plate, 1/8" to 1/2". These plates came in 11", 9", and 7" widths, and cut to length. They were pre-punched, alt. holes at appropriate spacing.

The plate was then sandwiched between the appropriate 2X's. It's called a flitch beam, and has a rational design approach established using the "n" calculation (Es/Ew) similar to steel/ concrete composite, and as described by someone earlier.

>>> "Bill Allen" <t.w.allen@cox.net> 8/7/2007 7:12 AM >>>

I would use the steel section to take the entire load regardless whether you remove the bottom laminations or not. Load sharing between the GLB and steel section is questionable (due to creep and other factors), especially when using through bolts with even 1/16" oversized holes. Bob Powell (and others) has made a good living providing expert witness testimony in cases such as yours. Is your E&O policy current?

When you say "I want to use a C10x15.3", why would you want to use something that doesn't work? Are you trying to stay with a 10" depth? If so, use a MC. You can go all the way up to an MC 10x41.3. If that still doesn't work, use a built-up section. Jack the (E) GLB to take out all of the deflection, attach the new steel section to the GLB (using the GLB only to transfer the load) then remove the shoring.

T. William (Bill) Allen, S.E.

ALLEN DESIGNS

Consulting Structural Engineers
V (949) 248-8588 • F(949) 209-2509

-----Original Message-----
From: Bill Cain [mailto:bcainse@aol.com]
Sent: Thursday, August 02, 2007 6:22 PM
To: seaint@seaint.org
Subject: Re: Glu-Lam beam with Steel Channel calc

If you are removing the tension lams, Like Tarek, I'd make sure the steel could take the entire load. You would need to consult with APA or AITC to determine new allowable stresses for the GLB with the lams removed if you want it to take anything. It is difficult to get a good connection between the two materials.Regards,

Bill Cain, SE

Berkeley CA

-----Original Message-----
From: Tarek Mokhtar <tarooky@earthlink.net>
To: seaint@seaint.org
Sent: Thu, 2 Aug 2007 11:42 am
Subject: RE: Glu-Lam beam with Steel Channel calc

Erik,

There are a few issues with transferring the load, bolting, stiffness, etc from the glb to the steel section, and it also appears that you are removing some of the high strength laminations from the bottom of the beam,

I would design the new steel section to handle ALL the load, but that's just me

Tarek Mokhtar, SE

Laguna Beach

When you say "taking off a few laminations" it sounds to me as if you are proposing cutting off the lower portion of the existing glulam beam, is that correct? If so, you'll want to be sure to take into account the new grade of lumber that you have at the tension face of your beam when figuring F'b.

-----Original Message-----
From: erik gibbs [mailto:erik.gibbs@gmail.com]
Sent: Thursday, August 02, 2007 10:59 AM
To: seaint@seaint.org
Subject: Glu-Lam beam with Steel Channel calc

I previously posted a question about a plan check question for out of plane loading on a shearwall and to all that posted I thank you, it cleared up any questions that I had.

Now to my new question. I have an existing 5-1/8" x 15" Glu Lam beam that spans 21'-6" in a garage of an existing residence. The owner wants to turn this into a flush beam, which means taking a few laminations off and sistering a new beam next to the GLB. I want to use a steel channel, C10x15.3, but this section, when checked in bending fails under the full load. Also the GLB fails under the full load when checked by itself. My question is how would you check/calc both beams in a rational method, instead of just saying that the GLB takes 50% of the load and the steel beam takes the other 50%?

Thanks

Erik Gibbs

--

Tarek Mokhtar, SE
TMM Structural Engineers, Inc
31645 S. Coast Hwy
Laguna Beach, CA., 92651
949-499-6254
949-499-2777 Fax

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Re: Glu-Lam beam with Steel Channel calc

Doesn't the channel need to be supported at the ends now and not the GLB, or some method of transferring the load back to the GLB at the support.

Joe Grill

----- Original Message -----

From: Bill Allen

To: seaint@seaint.org

Sent: Tuesday, August 07, 2007 5:12 AM

Subject: RE: Glu-Lam beam with Steel Channel calc
I would use the steel section to take the entire load regardless whether you remove the bottom laminations or not. Load sharing between the GLB and steel section is questionable (due to creep and other factors), especially when using through bolts with even 1/16" oversized holes. Bob Powell (and others) has made a good living providing expert witness testimony in cases such as yours. Is your E&O policy current?

When you say "I want to use a C10x15.3", why would you want to use something that doesn't work? Are you trying to stay with a 10" depth? If so, use a MC. You can go all the way up to an MC 10x41.3. If that still doesn't work, use a built-up section. Jack the (E) GLB to take out all of the deflection, attach the new steel section to the GLB (using the GLB only to transfer the load) then remove the shoring.

T. William (Bill) Allen, S.E.

ALLEN DESIGNS

Consulting Structural Engineers
V (949) 248-8588 • F(949) 209-2509
-----Original Message-----
From: Bill Cain [mailto:bcainse@aol.com]
Sent: Thursday, August 02, 2007 6:22 PM
To: seaint@seaint.org
Subject: Re: Glu-Lam beam with Steel Channel calc

If you are removing the tension lams, Like Tarek, I'd make sure the steel could take the entire load. You would need to consult with APA or AITC to determine new allowable stresses for the GLB with the lams removed if you want it to take anything. It is difficult to get a good connection between the two materials.Regards,

Bill Cain, SE

Berkeley CA

-----Original Message-----
From: Tarek Mokhtar <tarooky@earthlink.net>
To: seaint@seaint.org
Sent: Thu, 2 Aug 2007 11:42 am
Subject: RE: Glu-Lam beam with Steel Channel calc
Erik,

There are a few issues with transferring the load, bolting, stiffness, etc from the glb to the steel section, and it also appears that you are removing some of the high strength laminations from the bottom of the beam,

I would design the new steel section to handle ALL the load, but that's just me

Tarek Mokhtar, SE

Laguna Beach

When you say "taking off a few laminations" it sounds to me as if you are proposing cutting off the lower portion of the existing glulam beam, is that correct? If so, you'll want to be sure to take into account the new grade of lumber that you have at the tension face of your beam when figuring F'b.

-----Original Message-----
From: erik gibbs [mailto:erik.gibbs@gmail.com]
Sent: Thursday, August 02, 2007 10:59 AM
To: seaint@seaint.org
Subject: Glu-Lam beam with Steel Channel calc

I previously posted a question about a plan check question for out of plane loading on a shearwall and to all that posted I thank you, it cleared up any questions that I had.

Now to my new question. I have an existing 5-1/8" x 15" Glu Lam beam that spans 21'-6" in a garage of an existing residence. The owner wants to turn this into a flush beam, which means taking a few laminations off and sistering a new beam next to the GLB. I want to use a steel channel, C10x15.3, but this section, when checked in bending fails under the full load. Also the GLB fails under the full load when checked by itself. My question is how would you check/calc both beams in a rational method, instead of just saying that the GLB takes 50% of the load and the steel beam takes the other 50%?

Thanks

Erik Gibbs
-- 
Tarek Mokhtar, SE
TMM Structural Engineers, Inc
31645 S. Coast Hwy
Laguna Beach, CA., 92651
949-499-6254
949-499-2777 Fax
AOL now offers free email to everyone. Find out more about what's free from AOL at AOL.com.