Structural Engineering: 2009-09-06

Saturday, September 12, 2009

Re: Cold Formed Steel

Gary,
A couple questions ...
1) who is the girt manufacturer and is it part of a "system" assembly;
2) is the channel meant to have curved surfaces or is it damaged;
3) have they actually shipped the correct materials;
4) does the supplier/manufacturer claim that the materials are correct and
within tolerance?

Sometimes curved surfaces are used to resolve all possible tangential angles
in the assembly: field variances, sloping interfaces, etc. It is also
possible that they weren't loaded on the truck properly or stored at site
properly and got squashed.

I believe that you are describing a 3" deep girt which is very light. This
must be very small spans or spacing.

There are tolerances for square formed channels (e.g. 90 deg corners) in
S136 and A660 that could be extrapolated to your case. I will send a copy of
the tolerances - contact me directly if you need it sooner than Monday
morning.

Regards
Paul
--
Paul Ransom, P.Eng.
ph 905 639-9628
fax 905 639-3866
ad026@hwcn.org

> From: "Gary L. Hodgson and Assoc." <design@hodgsoneng.ca>

> List,
> I have a job where cold-formed steel girts were used. The girts are a
> channel shape 3"x3" x0.074. The web and flanges are curved rather than
> flat, approximately 5/16" out of flat from one corner to the other. My
> customer is claiming that the wall cladding cannot be properly installed
> and I agree with him. I checked the North American Specification for
> Cold Rolled Steel 2001 but it does not contain any information on
> tolerances. Does anybody out there have any information on acceptable
> tolerances? TIA
> Gary

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Re: Cold Formed Steel

Thanks, Jay.
Just what I was looking for.
Gary

Jay Parr wrote:

Gary,

The AISI Product Data standard (AISI S201-07) requires that structural members meet the tolerance requirements of ASTM C955. The tolerance for a crown in the web or flanges is plus or minus 1/16". Hope this helps.

Jay Parr

Friday, September 11, 2009

Cross-grain Compressive Strain in Timber

I'm doing some casual analysis of metal support plates bolted to the side
of, say, a built-up 2x10 beam. If I try to use, eg, concrete section
analogy, I need to know a strain modulus for the timber (yeah, I know timber
doesn't behave like concrete). What cross grain modulus might I use?

Eg: conc = .0035, masonry = .003, steel = .002 I could see a virtual SE =
.007? I know that ASTM tests to develop bearing stresses and lengths invoke
the concept of strain = del L/L and I also am aware that the value must be
very variable espec with MC.

Again, if I use strain compatibility analysis, I have to know what value to
use for the strain modulus .

Thanks

Thor A. Tandy P.Eng, C.Eng, Struct.Eng, MIStructE
Victoria, BC
Tel: (250) 382-9115

hst_ngc4414_9925Please consider the environment before printing out this
e-mail

RE: high uplift base plate connections

That is an option I am looking into. Thank you.

Can you put a plate on the bottom of the grade beam so threaded bars go through the beam?

Salvador Dorado wrote:

Andrew,

I am mostly concerned with the breakout of the concrete, I have designed the base plate using stiffeners to decrease the bending on the plate itself, and obtain a fair thickness for the plate. However, the concrete is what I am having trouble getting to work in breakout. Unfortunately, I don’t have much room to widen the concrete grade beam…any thoughts?

Re: high uplift base plate connections

Can you put a plate on the bottom of the grade beam so threaded bars go through the beam?

Salvador Dorado wrote:

Andrew,

I am mostly concerned with the breakout of the concrete, I have designed the base plate using stiffeners to decrease the bending on the plate itself, and obtain a fair thickness for the plate. However, the concrete is what I am having trouble getting to work in breakout. Unfortunately, I don’t have much room to widen the concrete grade beam…any thoughts?

RE: A Useful Resource for PDH and CEU Credits

AITC has some online continuing education courses for glulam design
available through Wiley Continuing Professional Education
(www.wileycpe.com). Two one-hour modules covering fundamentals of wood
and glulam are available now. Another four courses covering design of
glulam beams, columns, and beam-columns should be available in October.

Jeffrey D. Linville
Director, Technical Services
American Institute of Timber Construction
(303)792-9559
linville@aitc-glulam.org

-----Original Message-----
From: Gordon Goodell [mailto:GordonGoodell@harmonydesigninc.com]
Sent: Friday, September 11, 2009 10:40 AM
To: seaint@seaint.org
Subject: RE: A Useful Resource for PDH and CEU Credits

Very nice resource, Jeremy. Thank you.

Gordon Goodell

Jeremy White wrote on Friday, September 11, 2009 10:11 AM:

<<A colleague of mine has created the following website:

PDHonthecheap.com>>

Re: Concrete Cure Time

Currently in southern CA slag is not available at an economically
feasible cost; FA is your only realistic option for recycled cementious
material replacement. High fly ash (greater than 40% replacement) is
not common either down here but it does seem to be picking up steam and
is more common in other parts of the world. I have done a little
research on the subject and do not recall any info regarding Em:f'c
differences in HFA vs 100% PC concrete. Doesn't mean there isn't one as
I am pretty much a novice in this area. If you are interested send me
an email and I will forward some of the documents and information I have
complied from the internet and other sources on high fly ash in general
as well as specific to southern CA.

Donny Harris, SE
California

Contractors will want you to get rid of the flyash completely and use
Slag.
The concrete apparently stays too wet for too long using high amounts
fly
ash.

-gm

On Thu, Sep 10, 2009 at 3:17 PM, <ASLCSE@aol.com> wrote:

> Great input, Harold. I have a related (I think) question regarding
> strength.
> A couple of days ago I went to a GREEN seminar. The speaker told us
that
> using a great amount of fly ash (replacing part of the PC) will
greatly
> reduce the CO2 into the air. I know that fly ash will also act as a
retar=
der
> (for the concrete strength). Question: Is the strength vs. the modulus
of
> elasticity curve the same as for using portland cement ?
> Thanks
> Antonio S. "Tony" Luisoni
> Consulting SE
> Granada Hills, CA

RE: concrete anchorage using appendix D of ACI-318-05

I have had the same problem as you described. My solution was to provide concrete reinforcement that is developed within the shear cone of the bolt group. This will transfer the tension force to the reinforcement. The key is to make sure the bar is developed and hooked within the shear cone.

Andre Sidler
Quantum Consulting Engineers
Seattle, WA

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15 Message:0015 15
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From: "Salvador Dorado" <sdorado@tbengineeringinc.com>
To: <seaint@seaint.org>
Subject: RE: concrete anchorage using appendix D of ACI-318-05

This is a multi-part message in MIME format.

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Hi,

Is anyone able to provide me with a sample calc for anchoring a
brace frame base plate to concrete. The uplift load I have is 335 kips. I
have worked on it but the size I obtain for the grade beam size to resist
the uplift load in breakout seems quite high. I have to use a 6'6" grade
beam and embed the anchors at least 60 inches in order to get it to work.
Any feedback would be great. Thank you.

Re: concrete anchorage using appendix D of ACI-318-05

Salvador,

As I read this you have 335 kips of uplift; concrete weighs about 4 kips per cubic yard; so you need 335/4 = 84 cubic yards of dead weight concrete to hold this down, not including a safety factor.

Am I missing something?

Regards,

H. Daryl Richardson

----- Original Message -----

From: Salvador Dorado

To: seaint@seaint.org

Sent: Thursday, September 10, 2009 3:33 PM

Subject: RE: concrete anchorage using appendix D of ACI-318-05

Hi,

Is anyone able to provide me with a sample calc for anchoring a brace frame base plate to concrete. The uplift load I have is 335 kips. I have worked on it but the size I obtain for the grade beam size to resist the uplift load in breakout seems quite high. I have to use a 6'6" grade beam and embed the anchors at least 60 inches in order to get it to work. Any feedback would be great. Thank you.

RE: A Useful Resource for PDH and CEU Credits

Very nice resource, Jeremy. Thank you.

Gordon Goodell

Jeremy White wrote on Friday, September 11, 2009 10:11 AM:

<<A colleague of mine has created the following website:

PDHonthecheap.com>>

Re: Steel Building Design

If you are looking in Table 15.2 you are not dealing with a building.

Joe Venuti
Johnson & Nielsen Associates
Palm Springs, CA

In a message dated 9/11/2009 9:11:26 A.M. Pacific Daylight Time, cobdel@hotmail.com writes:

Table 15.4-1 for OMF with R=2.5 with increased height point to detail requirement AISC 341 (Seismic Steel provisions). In the AISC 341 state that this not applicable to buildings with R equal or less than 3. Therefore, the detail requirement default to AISC 360 (Standard Steel specifiactions 13th edition). It seems to be a disagreement between the AISC and the ASCE.

Does anybody has additional information to this matter?

Thanks

Get back to school stuff for them and cashback for you. Try Bing now. =

RE: high uplift base plate connections

Sorry I haven’t been following this thread but can you attach down the sides of your concrete with side plates. I’ve done that when my wall is too thin for the anchor design prescription.

Thor A. Tandy P.Eng, C.Eng, Struct.Eng, MIStructE
Victoria, BC
Canada

From: Salvador Dorado [mailto:sdorado@tbengineeringinc.com]
Sent: Friday, September 11, 2009 9:22 AM
To: seaint@seaint.org
Subject: RE: high uplift base plate connections

Andrew,

RE: high uplift base plate connections

Andrew,

A Useful Resource for PDH and CEU Credits

A colleague of mine has created the following website:

PDHonthecheap.com

On this website you can search for or submit free or low cost events
that help engineers fulfill their PDH/CEU requirements.

It is in its beta phase and we are still brainstorming on additional
features that might make it more useful so if you have any comments
feel free to contact me.

Regards,
Jeremy White

re: high uplift base plate connections

What others have mentioned previously on this list, and what I have done in the past, is to rely on a different mechanism to develop your uplift force outside of ACI Appendix D. This can include many methods such as:

- lapping threaded rods attached to the base plate with rebar in the footing, developing tension via lap splices, welds, or mechanical means

-using weldable rebar attached to the base plate/columns and developing it via lap splices with footing rebar

-embedding a steel plate or other structural member into the footing and relying on upward punching shear and reinforcement in the shear zone of the footing to transfer the uplift force. The steel plate would be attached to the base plate via threaded rods or other mechanical means

-helical or auger piles with all-thread connections to the base plate; use of lap splices in the case of auger piles

-also, I forgot the proper name, but with very high uplift forces local bending in the base plate is likely a concern. You may have to connect the threaded rods or whatever your anchorage method is directly to your column with a reinforced local connection. That way the base plate is only in compression and is not used in resisting the uplift.

I am sure others have some I may have left out. At extremely high pullout forces it becomes almost a bridge-dead-man anchor situation, and some of those methods of tensile anchorage may be studied. Or something from the post-tension industry. Way outside my field of expertise…

HTH,

Andrew Kester, PE

Orlando, FL

Steel Building Design

Table 15.4-1 for OMF with R=2.5 with increased height point to detail requirement AISC 341 (Seismic Steel provisions). In the AISC 341 state that this not applicable to buildings with R equal or less than 3. Therefore, the detail requirement default to AISC 360 (Standard Steel specifiactions 13th edition). It seems to be a disagreement between the AISC and the ASCE.

Does anybody has additional information to this matter?

Thanks

Get back to school stuff for them and cashback for you. Try Bing now.

Re: California Building Code

Scott,

As the late Ed McMahon would bellow to the man behind the desk, "YOU ARE CORRECT, SIR!"

On Thu, Sep 10, 2009 at 10:12 PM, Scott Maxwell <smaxwell@umich.edu> wrote:

When you say "California Special CE exam", I am assuming you mean the seismic exam that is required to get your PE in California as opposed to the exam to get your SE license. If so, then I took it and passed it with just a regular UBC (this was before the switch to the IBC). From what I recall, the exam does not even touch the issue of the "A" chapters of the CBC (I suspect because technically only SEs use those chapters since schools and hospitals are the only things that require the "A" chapters to my knowledge and those must be done by SEs).

HTH,

Regards,

Scott
Adrian, MI

On 9/9/09 5:34 PM, "Padmanabhan Rajendran" <prajendran@ymail.com> wrote:

I am preparing for the California Special CE exam.

The California Building Code (CBC) has a Chapter 16A following Chapter 16. each of the chapters are 26 pages long. I have not compared paragraph by paragraph, but, randomly looked at the basic load combinations 1605.2.1 and 1605A.2.1. The only difference was the addition of paragraph of 1605A.2.1.1 in chapter 16A.

Considering that CBC is an adaptation of IBC 2006, it would have been simpler to keep one chapter with all required changes instead of retaining the original IBC chapter with footnotes and adding another chapter with identical paragraphs.

Am I missing something? Perhaps, the list members from California could clarify.

Thanks.

Rajendran

--
David Topete, SE

Re: California Building Code

Thanks, Scott. That is a valuable information..

Rajendran

--- On Fri, 9/11/09, Scott Maxwell <smaxwell@umich.edu> wrote:

From: Scott Maxwell <smaxwell@umich.edu>
Subject: Re: California Building Code
To: seaint@seaint.org
Date: Friday, September 11, 2009, 5:12 AM

Re: California Building Code When you say "California Special CE exam", I am assuming you mean the seismic exam that is required to get your PE in California as opposed to the exam to get your SE license. If so, then I took it and passed it with just a regular UBC (this was before the switch to the IBC). From what I recall, the exam does not even touch the issue of the "A" chapters of the CBC (I suspect because technically only SEs use those chapters since schools and hospitals are the only things that require the "A" chapters to my knowledge and those must be done by SEs).

HTH,

Regards,

Scott
Adrian, MI

On 9/9/09 5:34 PM, "Padmanabhan Rajendran" <prajendran@ymail.com> wrote:

I am preparing for the California Special CE exam.

The California Building Code (CBC) has a Chapter 16A following Chapter 16. each of the chapters are 26 pages long. I have not compared paragraph by paragraph, but, randomly looked at the basic load combinations 1605.2.1 and 1605A.2.1. The only difference was the addition of paragraph of 1605A.2.1.1 in chapter 16A.

Considering that CBC is an adaptation of IBC 2006, it would have been simpler to keep one chapter with all required changes instead of retaining the original IBC chapter with footnotes and adding another chapter with identical paragraphs.

Am I missing something? Perhaps, the list members from California could clarify.

Thanks.

Rajendran

RE: Seismic Joints

My recollection is that the removal of the SRSS method from the 2006 IBC was unintentional—it was accidentally deleted when the bulk of the seismic provisions (except for the SDC determination) were pulled from the IBC and replaced with a direct reference to ASCE 7-05. The proposal to restore it came from SEAOC, which would certainly indicate it never “went out of vogue”, regardless of what was or wasn’t in ASCE 7-05 or the 2006 IBC.

Gary

Gary J. Ehrlich, PE
Program Manager, Structural Codes & Standards
National Association of Home Builders (NAHB)
1201 15th Street, NW, Washington, DC 20005
ph: 202-266-8545 or 800-368-5242 x8545
fax: 202-266-8369
gehrlich@nahb.com

Attend the 2010 International Builders' Show
January 19-22, 2010, Las Vegas, NV
www.buildersshow.com

www.builderbooks.com

www.housingeconomics.com

From: Doug Mayer [mailto:doug.mayer@taylorteter.com]
Sent: Thursday, September 10, 2009 5:57 PM
To: seaint@seaint.org
Subject: RE: Seismic Joints

Gerard,

As Gary mentioned, the SRSS for building separation will be back in the 2009 IBC and hopefully in the CBC soon thereafter. Luckily, I found out about this early on when I had to deal with a seismic separation and I was told on good authority that it should still be calculated using SRSS. I’ve done a couple of buildings with separation since then using SRSS and I haven’t been called on it yet.

Doug Mayer, SE

Structural Engineer

From: Ehrlich, Gary [mailto:gehrlich@nahb.com]
Sent: Thursday, September 10, 2009 10:48 AM
To: seaint@seaint.org
Subject: RE: Seismic Joints

Gerard,

Don’t know if this helps any, but the SRSS method was reintroduced into the 2009 IBC. New section 1613.6.7. I believe the SRSS method will also be in ASCE 7-10.

Gary

Attend the 2010 International Builders' Show
January 19-22, 2010, Las Vegas, NV
www.buildersshow.com

www.builderbooks.com

www.housingeconomics.com

From: Gerard Madden, SE [mailto:gmse4603@gmail.com]
Sent: Thursday, September 10, 2009 1:09 PM
To: seaint@seaint.org
Subject: Seismic Joints

The ASCE 7 makes a very vague statement about building separations in section 12.12.3

In the good old UBC, we were using SRSS to determine the gap to avoid pounding.

The IBC 2006 Design guide volume I indicates that this gap should be the SUM of the inelastic deflections.

So, say for a building with a drift limit of 1% (an essential facility) with 50 feet to the roof, the gap would need to be 8.5" under the SRSS method. Under the IBC example it would need to be 12"

If I had a regular occupancy building and could use 2.5% drift, then I would need a 30" joint

50 feet is about a 4 story building....are we really going into 24 " seismic gaps for 8-9 story buildings now?

Feedback appreciated.

-gm

Re: Cold Formed Steel

Gary,

The AISI Product Data standard (AISI S201-07) requires that structural members meet the tolerance requirements of ASTM C955. The tolerance for a crown in the web or flanges is plus or minus 1/16". Hope this helps.

Jay Parr

Re: Cold Formed Steel

Thanks, Conrad
This is better than what I had up to now which is zilch.
Gary

Conrad Harrison wrote:
> Gary,
>
> Not aware of any tolerances for cold-formed steel. Sections can be made to
> any dimension and shape, and for a multitude of applications, including
> deliberately having curved flanges. So required tolerances would be
> dependent on the application, some manufacturers do indicate what they can
> achieve.
>
> There is some guidance for acceptable tolerances for steel house framing,
> which is mainly channel like sections. The NASH draft specification, says:
>
> D1.1.1 Cold-formed sections
> a) Material thickness shall conform to AS1397.
> b) Tolerances of sections, assuming design thickness, shall be determined
> such that the relevant actual sectional properties are not more than ±5%
> from the design section properties.
> c) Tolerances appropriate for particular sections shall be specified to
> comply with the above.
>
> AS1397 is just a standard for the production of steel coil and strip,
> probably derived from an equivalent ASTM specification.
>
> One manufacturer of channel like sections indicates ±2 mm on depth and ±2 mm
> on flange width. Assuming envelope method, then any actual shape which fits
> between the envelopes for the maximum material condition (MMC) and the least
> material condition (LMC) would be acceptable. Which suggests a maximum rise
> or fall for curvature of about 1mm. Much less than the 5/16" (7.9mm).
>
> Another guide is simply to use tolerances given in codes for hot rolled
> sections. These tolerances should be based on requirements for suitability
> for inclusion in finished construction, rather than based on expectations
> from manufacturing process.
>
> If the sections are meant to be flat and square then doesn't sound good,
> both roll forming and folding should achieve better than described.
>
>
> Regards
> Conrad Harrison
> B.Tech (mfg & mech), MIIE, gradTIEAust
> mailto:sch.tectonic@bigpond.com
> Adelaide
> South Australia
>
>
>
> ******* ****** ******* ******** ******* ******* ******* ***
> * Read list FAQ at: http://www.seaint.org/list_FAQ.asp
> *
> * This email was sent to you via Structural Engineers
> * Association of Southern California (SEAOSC) server. To
> * subscribe (no fee) or UnSubscribe, please go to:
> *
> * http://www.seaint.org/sealist1.asp
> *
> * Questions to seaint-ad@seaint.org. Remember, any email you
> * send to the list is public domain and may be re-posted
> * without your permission. Make sure you visit our web
> * site at: http://www.seaint.org
> ******* ****** ****** ****** ******* ****** ****** ********
>
>

Thursday, September 10, 2009

Re: California Building Code

When you say “California Special CE exam”, I am assuming you mean the seismic exam that is required to get your PE in California as opposed to the exam to get your SE license. If so, then I took it and passed it with just a regular UBC (this was before the switch to the IBC). From what I recall, the exam does not even touch the issue of the “A” chapters of the CBC (I suspect because technically only SEs use those chapters since schools and hospitals are the only things that require the “A” chapters to my knowledge and those must be done by SEs).

HTH,

Regards,

Scott
Adrian, MI

On 9/9/09 5:34 PM, "Padmanabhan Rajendran" <prajendran@ymail.com> wrote:

I am preparing for the California Special CE exam.

The California Building Code (CBC) has a Chapter 16A following Chapter 16. each of the chapters are 26 pages long. I have not compared paragraph by paragraph, but, randomly looked at the basic load combinations 1605.2.1 and 1605A.2.1. The only difference was the addition of paragraph of 1605A.2.1.1 in chapter 16A.

Considering that CBC is an adaptation of IBC 2006, it would have been simpler to keep one chapter with all required changes instead of retaining the original IBC chapter with footnotes and adding another chapter with identical paragraphs.

Am I missing something? Perhaps, the list members from California could clarify.

Thanks.

Rajendran

RE: Cold Formed Steel

Gary,

Sorry. Correction. That would be a maximum rise or fall of 2mm (0.0787
inches). (1mm below nominal and 1mm above nominal for each flange. With both
flanges splayed out, get total increase in depth of 2mm at open face, and
2mm decrease in depth at web. The profile fitting between the extreme MMC
and LMC envelopes)

Regards
Conrad Harrison
B.Tech (mfg & mech), MIIE, gradTIEAust
mailto:sch.tectonic@bigpond.com
Adelaide
South Australia

RE: Cold Formed Steel

Gary,

Not aware of any tolerances for cold-formed steel. Sections can be made to
any dimension and shape, and for a multitude of applications, including
deliberately having curved flanges. So required tolerances would be
dependent on the application, some manufacturers do indicate what they can
achieve.

There is some guidance for acceptable tolerances for steel house framing,
which is mainly channel like sections. The NASH draft specification, says:

D1.1.1 Cold-formed sections
a) Material thickness shall conform to AS1397.
b) Tolerances of sections, assuming design thickness, shall be determined
such that the relevant actual sectional properties are not more than ±5%
from the design section properties.
c) Tolerances appropriate for particular sections shall be specified to
comply with the above.

AS1397 is just a standard for the production of steel coil and strip,
probably derived from an equivalent ASTM specification.

One manufacturer of channel like sections indicates ±2 mm on depth and ±2 mm
on flange width. Assuming envelope method, then any actual shape which fits
between the envelopes for the maximum material condition (MMC) and the least
material condition (LMC) would be acceptable. Which suggests a maximum rise
or fall for curvature of about 1mm. Much less than the 5/16" (7.9mm).

Another guide is simply to use tolerances given in codes for hot rolled
sections. These tolerances should be based on requirements for suitability
for inclusion in finished construction, rather than based on expectations
from manufacturing process.

If the sections are meant to be flat and square then doesn't sound good,
both roll forming and folding should achieve better than described.

Regards
Conrad Harrison
B.Tech (mfg & mech), MIIE, gradTIEAust
mailto:sch.tectonic@bigpond.com
Adelaide
South Australia

Re: concrete anchorage using appendix D of ACI-318-05

I think I have a PDF of the 2nd edition of this design guide. If I can find it, I will email it to you.

Salvador Dorado wrote:

Could anyone provide me with a scanned copy. Or are these design guides sold? If it were available?

Regards,

Salvador Dorado

Project Engineer

Engineering, Inc.

consulting structural engineers

4344 Latham St., Suite 100

Riverside, CA 92501-1773

P: (951) 684-6200

F: (951) 684-6226

From: Drew Morris [mailto:dmorris@bbfm.com]
Sent: Thursday, September 10, 2009 4:33 PM
To: seaint@seaint.org
Subject: Re: concrete anchorage using appendix D of ACI-318-05

Design Guide 1: Base Plate and Anchor Rod Design (2nd Edition) has been taken off the AISC website. AISC now says:

"We are currently working on the third edition of Design Guide 1, Base Plate and Anchor Rod Design.
The Second Edition of this design guide is no longer available."

Gerard Madden, SE wrote:

Design Guide 1

On Thu, Sep 10, 2009 at 3:35 PM, Salvador Dorado <sdorado@tbengineeringinc.com> wrote:

Gerard, which AISC might I find a method in?

Use an AISC method.

On Thu, Sep 10, 2009 at 2:33 PM, Salvador Dorado <sdorado@tbengineeringinc.com> wrote:

Hi,

Is anyone able to provide me with a sample calc for anchoring a brace frame base plate to concrete. The uplift load I have is 335 kips. I have worked on it but the size I obtain for the grade beam size to resist the uplift load in breakout seems quite high. I have to use a 6’6” grade beam and embed the anchors at least 60 inches in order to get it to work. Any feedback would be great. Thank you.

Re: concrete anchorage using appendix D of ACI-318-05

I bought them in a packet of 15 or 16 design guides about 6 years ago. I think you can buy them individually and by PDF now....although Drew indicates they've yanked this one for the time being.

-gm

On Thu, Sep 10, 2009 at 4:59 PM, Salvador Dorado <sdorado@tbengineeringinc.com> wrote:

Could anyone provide me with a scanned copy. Or are these design guides sold? If it were available?

Regards,

Salvador Dorado

Project Engineer

Engineering, Inc.

consulting structural engineers

4344 Latham St., Suite 100

Riverside, CA 92501-1773

P: (951) 684-6200

F: (951) 684-6226

From: Drew Morris [mailto:dmorris@bbfm.com]
Sent: Thursday, September 10, 2009 4:33 PM
To: seaint@seaint.org
Subject: Re: concrete anchorage using appendix D of ACI-318-05

Design Guide 1: Base Plate and Anchor Rod Design (2nd Edition) has been taken off the AISC website. AISC now says:

"We are currently working on the third edition of Design Guide 1, Base Plate and Anchor Rod Design.
The Second Edition of this design guide is no longer available."

Gerard Madden, SE wrote:

Design Guide 1

On Thu, Sep 10, 2009 at 3:35 PM, Salvador Dorado <sdorado@tbengineeringinc.com> wrote:

Gerard, which AISC might I find a method in?

Use an AISC method.

On Thu, Sep 10, 2009 at 2:33 PM, Salvador Dorado <sdorado@tbengineeringinc.com> wrote:

Hi,

Is anyone able to provide me with a sample calc for anchoring a brace frame base plate to concrete. The uplift load I have is 335 kips. I have worked on it but the size I obtain for the grade beam size to resist the uplift load in breakout seems quite high. I have to use a 6'6" grade beam and embed the anchors at least 60 inches in order to get it to work. Any feedback would be great. Thank you.

RE: concrete anchorage using appendix D of ACI-318-05

Could anyone provide me with a scanned copy. Or are these design guides sold? If it were available?

Regards,

Salvador Dorado

Project Engineer

cid:image002.jpg@01C923CF.C1ACF5B0 Engineering, Inc.

consulting structural engineers

4344 Latham St., Suite 100

Riverside, CA 92501-1773

P: (951) 684-6200

F: (951) 684-6226

From: Drew Morris [mailto:dmorris@bbfm.com]
Sent: Thursday, September 10, 2009 4:33 PM
To: seaint@seaint.org
Subject: Re: concrete anchorage using appendix D of ACI-318-05

Design Guide 1: Base Plate and Anchor Rod Design (2nd Edition) has been taken off the AISC website. AISC now says:

"We are currently working on the third edition of Design Guide 1, Base Plate and Anchor Rod Design.
The Second Edition of this design guide is no longer available."

Gerard Madden, SE wrote:

Design Guide 1

On Thu, Sep 10, 2009 at 3:35 PM, Salvador Dorado <sdorado@tbengineeringinc.com> wrote:

Gerard, which AISC might I find a method in?

Use an AISC method.

On Thu, Sep 10, 2009 at 2:33 PM, Salvador Dorado <sdorado@tbengineeringinc.com> wrote:

Hi,

Is anyone able to provide me with a sample calc for anchoring a brace frame base plate to concrete. The uplift load I have is 335 kips. I have worked on it but the size I obtain for the grade beam size to resist the uplift load in breakout seems quite high. I have to use a 6’6” grade beam and embed the anchors at least 60 inches in order to get it to work. Any feedback would be great. Thank you.

Re: concrete anchorage using appendix D of ACI-318-05

Design Guide 1

On Thu, Sep 10, 2009 at 3:35 PM, Salvador Dorado <sdorado@tbengineeringinc.com> wrote:

Gerard, which AISC might I find a method in?

Use an AISC method.

On Thu, Sep 10, 2009 at 2:33 PM, Salvador Dorado <sdorado@tbengineeringinc.com> wrote:

Hi,

Is anyone able to provide me with a sample calc for anchoring a brace frame base plate to concrete. The uplift load I have is 335 kips. I have worked on it but the size I obtain for the grade beam size to resist the uplift load in breakout seems quite high. I have to use a 6'6" grade beam and embed the anchors at least 60 inches in order to get it to work. Any feedback would be great. Thank you.

Re: concrete anchorage using appendix D of ACI-318-05

Design Guide 1

On Thu, Sep 10, 2009 at 3:35 PM, Salvador Dorado <sdorado@tbengineeringinc.com> wrote:

Gerard, which AISC might I find a method in?

Use an AISC method.

On Thu, Sep 10, 2009 at 2:33 PM, Salvador Dorado <sdorado@tbengineeringinc.com> wrote:

Hi,

Is anyone able to provide me with a sample calc for anchoring a brace frame base plate to concrete. The uplift load I have is 335 kips. I have worked on it but the size I obtain for the grade beam size to resist the uplift load in breakout seems quite high. I have to use a 6'6" grade beam and embed the anchors at least 60 inches in order to get it to work. Any feedback would be great. Thank you.

Re: Concrete Cure Time

Contractors will want you to get rid of the flyash completely and use Slag. The concrete apparently stays too wet for too long using high amounts fly ash.

-gm

On Thu, Sep 10, 2009 at 3:17 PM, <ASLCSE@aol.com> wrote:

Great input, Harold. I have a related (I think) question regarding strength.

A couple of days ago I went to a GREEN seminar. The speaker told us that using a great amount of fly ash (replacing part of the PC) will greatly reduce the CO2 into the air. I know that fly ash will also act as a retarder (for the concrete strength). Question: Is the strength vs. the modulus of elasticity curve the same as for using portland cement ?

Thanks

Antonio S. "Tony" Luisoni

Consulting SE

Granada Hills, CA

In a message dated 9/2/2009 6:54:06 A.M. Pacific Daylight Time, spraguehope@hotmail.com writes:

Don't leave it arbitrary, or a rule of thumb. Calculations can be developed for concrete at various ages and temperatures.

There are many good guides out there. I would suggest that you look at the Concrete Manual, US Dept of Interior, Figure 10 as a guide. It is a graph showing concrete age on the horizontal axis and percent of strength on the vertical axis. The chart contains various temperatures of concrete so that you can anticipate strength gain at various times.

For example a 70 degree F ambient will have about 35% of the 28 day strength in 3 days. If your target is 1,500 psf for a slab on grade for construction, your 28 day target should be about 4,300 psi. In 7 days you will get about 65% of your 28 day target at 70 degrees.

If it is important, you can develop your mix for strengths at various times. The example cited above is subject to other aspects of the mix matrix like polycarboxolates and flyash to get that LEED point.

Regards, Harold Sprague

From: RichardC@lbbe.com
To: seaint@seaint.org
Date: Wed, 2 Sep 2009 08:15:53 -0400
Subject: RE: Concrete Cure Time

Thanks for everyone's input on this one. As far as I know the 3 day rule has just been something of a rule of thumb in this area. It appears to me that even at an early stage concrete will have enough strength to sustain a day or so worth of hollow block work – that's not really all that much pressure. I'm going to suggest that we allow the placement at 24 hours, but not allow the cells to be poured until a yet to be determined time frame.

Again, thanks.

From: Jim Getaz [mailto:jgetaz@shockeyprecast.com]
Sent: Wednesday, September 02, 2009 7:23 AM
To: seaint@seaint.org
Subject: Re: Concrete Cure Time

        Richard,

                I concur with Harold and Jordan. We transfer prestress at 12-16 hours, depending on how long the crew worked the day before. Our minimum is 3,500 psi, and can be as high as 4,200 psi with our normal mix. We break two cylinders (three, soon, I guess) for a full test before cutting strand. We used to use high-early, but cannot purchase it anymore. I doubt our normal mix would resemble the average slab-on-grade normal mix other than rocks, water and cement, but let them prove they can proceed. This is not your call.

        Jim Getaz

        Precast Concrete Engineer

With Windows Live, you can organize, edit, and share your photos. Click here. =

RE: concrete anchorage using appendix D of ACI-318-05

Gerard, which AISC might I find a method in?

Use an AISC method.

On Thu, Sep 10, 2009 at 2:33 PM, Salvador Dorado <sdorado@tbengineeringinc.com> wrote:

Hi,

Re: Concrete Cure Time

Great input, Harold. I have a related (I think) question regarding strength.

A couple of days ago I went to a GREEN seminar. The speaker told us that using a great amount of fly ash (replacing part of the PC) will greatly reduce the CO2 into the air. I know that fly ash will also act as a retarder (for the concrete strength). Question: Is the strength vs. the modulus of elasticity curve the same as for using portland cement ?

Thanks

Antonio S. "Tony" Luisoni

Consulting SE

Granada Hills, CA

In a message dated 9/2/2009 6:54:06 A.M. Pacific Daylight Time, spraguehope@hotmail.com writes:

Don't leave it arbitrary, or a rule of thumb. Calculations can be developed for concrete at various ages and temperatures.

There are many good guides out there. I would suggest that you look at the Concrete Manual, US Dept of Interior, Figure 10 as a guide. It is a graph showing concrete age on the horizontal axis and percent of strength on the vertical axis. The chart contains various temperatures of concrete so that you can anticipate strength gain at various times.

For example a 70 degree F ambient will have about 35% of the 28 day strength in 3 days. If your target is 1,500 psf for a slab on grade for construction, your 28 day target should be about 4,300 psi. In 7 days you will get about 65% of your 28 day target at 70 degrees.

If it is important, you can develop your mix for strengths at various times. The example cited above is subject to other aspects of the mix matrix like polycarboxolates and flyash to get that LEED point.

Regards, Harold Sprague

From: RichardC@lbbe.com
To: seaint@seaint.org
Date: Wed, 2 Sep 2009 08:15:53 -0400
Subject: RE: Concrete Cure Time

Thanks for everyone's input on this one. As far as I know the 3 day rule has just been something of a rule of thumb in this area. It appears to me that even at an early stage concrete will have enough strength to sustain a day or so worth of hollow block work – that's not really all that much pressure. I'm going to suggest that we allow the placement at 24 hours, but not allow the cells to be poured until a yet to be determined time frame.

Again, thanks.

From: Jim Getaz [mailto:jgetaz@shockeyprecast.com]
Sent: Wednesday, September 02, 2009 7:23 AM
To: seaint@seaint.org
Subject: Re: Concrete Cure Time

        Richard,

                I concur with Harold and Jordan. We transfer prestress at 12-16 hours, depending on how long the crew worked the day before. Our minimum is 3,500 psi, and can be as high as 4,200 psi with our normal mix. We break two cylinders (three, soon, I guess) for a full test before cutting strand. We used to use high-early, but cannot purchase it anymore. I doubt our normal mix would resemble the average slab-on-grade normal mix other than rocks, water and cement, but let them prove they can proceed. This is not your call.

        Jim Getaz

        Precast Concrete Engineer

With Windows Live, you can organize, edit, and share your photos. Click here. =

Re: Seismic Joints

Also,

I did find this which should help overriding the example I saw in the IBC seismic design guide volume I

http://www.iccsafe.org/cs/codes/errata/2006/IBC/2006_IBC_Stuctural_Vol_1_errata_1-3_Printing.pdf

-gm

On Thu, Sep 10, 2009 at 3:06 PM, Gerard Madden, SE <gmse4603@gmail.com> wrote:

Great, thanks Doug...good to know.

-gm

On Thu, Sep 10, 2009 at 2:56 PM, Doug Mayer <doug.mayer@taylorteter.com> wrote:

Gerard,

As Gary mentioned, the SRSS for building separation will be back in the 2009 IBC and hopefully in the CBC soon thereafter. Luckily, I found out about this early on when I had to deal with a seismic separation and I was told on good authority that it should still be calculated using SRSS. I've done a couple of buildings with separation since then using SRSS and I haven't been called on it yet.

Doug Mayer, SE

Structural Engineer

From: Ehrlich, Gary [mailto:gehrlich@nahb.com]
Sent: Thursday, September 10, 2009 10:48 AM
To: seaint@seaint.org
Subject: RE: Seismic Joints

Gerard,

Don't know if this helps any, but the SRSS method was reintroduced into the 2009 IBC. New section 1613.6.7. I believe the SRSS method will also be in ASCE 7-10.

Gary

Gary J. Ehrlich, PE
Program Manager, Structural Codes & Standards
National Association of Home Builders (NAHB)
1201 15th Street, NW, Washington, DC 20005
ph: 202-266-8545 or 800-368-5242 x8545
fax: 202-266-8369
gehrlich@nahb.com

Attend the 2010 International Builders' Show
January 19-22, 2010, Las Vegas, NV
www.buildersshow.com

www.builderbooks.com

www.housingeconomics.com

From: Gerard Madden, SE [mailto:gmse4603@gmail.com]
Sent: Thursday, September 10, 2009 1:09 PM
To: seaint@seaint.org
Subject: Seismic Joints

The ASCE 7 makes a very vague statement about building separations in section 12.12.3

In the good old UBC, we were using SRSS to determine the gap to avoid pounding.

The IBC 2006 Design guide volume I indicates that this gap should be the SUM of the inelastic deflections.

So, say for a building with a drift limit of 1% (an essential facility) with 50 feet to the roof, the gap would need to be 8.5" under the SRSS method. Under the IBC example it would need to be 12"

If I had a regular occupancy building and could use 2.5% drift, then I would need a 30" joint

50 feet is about a 4 story building....are we really going into 24 " seismic gaps for 8-9 story buildings now?

Feedback appreciated.

-gm

Re: Seismic Joints

Great, thanks Doug...good to know.

-gm

On Thu, Sep 10, 2009 at 2:56 PM, Doug Mayer <doug.mayer@taylorteter.com> wrote:

Gerard,

As Gary mentioned, the SRSS for building separation will be back in the 2009 IBC and hopefully in the CBC soon thereafter. Luckily, I found out about this early on when I had to deal with a seismic separation and I was told on good authority that it should still be calculated using SRSS. I've done a couple of buildings with separation since then using SRSS and I haven't been called on it yet.

Doug Mayer, SE

Structural Engineer

From: Ehrlich, Gary [mailto:gehrlich@nahb.com]
Sent: Thursday, September 10, 2009 10:48 AM
To: seaint@seaint.org
Subject: RE: Seismic Joints

Gerard,

Don't know if this helps any, but the SRSS method was reintroduced into the 2009 IBC. New section 1613.6.7. I believe the SRSS method will also be in ASCE 7-10.

Gary

Gary J. Ehrlich, PE
Program Manager, Structural Codes & Standards
National Association of Home Builders (NAHB)
1201 15th Street, NW, Washington, DC 20005
ph: 202-266-8545 or 800-368-5242 x8545
fax: 202-266-8369
gehrlich@nahb.com

Attend the 2010 International Builders' Show
January 19-22, 2010, Las Vegas, NV
www.buildersshow.com

www.builderbooks.com

www.housingeconomics.com

From: Gerard Madden, SE [mailto:gmse4603@gmail.com]
Sent: Thursday, September 10, 2009 1:09 PM
To: seaint@seaint.org
Subject: Seismic Joints

The ASCE 7 makes a very vague statement about building separations in section 12.12.3

In the good old UBC, we were using SRSS to determine the gap to avoid pounding.

The IBC 2006 Design guide volume I indicates that this gap should be the SUM of the inelastic deflections.

So, say for a building with a drift limit of 1% (an essential facility) with 50 feet to the roof, the gap would need to be 8.5" under the SRSS method. Under the IBC example it would need to be 12"

If I had a regular occupancy building and could use 2.5% drift, then I would need a 30" joint

50 feet is about a 4 story building....are we really going into 24 " seismic gaps for 8-9 story buildings now?

Feedback appreciated.

-gm